MATERIAL PROPERTffiS AND CORROSION CONDITION OF A TWENTY-YEAR-OLD PRESTRESSED BRIDGE GIRDER FINAL REPORT

Transcription

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2 MATERIAL PROPERTffiS AND CORROSION CONDITION OF A TWENTY-YEAR-OLD PRESTRESSED BRIDGE GIRDER FINAL REPORT Prepared by Florie B. Coggins Catherine W. French University of Minnesota. Department of Civil and Mineral Engineering May 1992 Submitted to Minnesota. Department of Transportation Office of Research Administration 2 Ford Building, 117 University Avenue St. Paul, MN The opinions, findings and conclusions expressed in this publication are those of the authors and not necessarily those of the sponsors.

3 ABSTRACT This report presents the results of a series of nondestructive tests to estimate the strength of concrete in a twenty-year-old prestressed bridge girder removed from an interstate overpass. Nondestructive test results were compared with compressive strength tests of cores obtained from both the girder tested nondestructively and a second girder from the bridge. In addition, the chloride ion concentration of the concrete was analyzed to determine whether corrosion of the bridge reinforcing steel may have occurred. The nondestructive tests performed included rebound hammer, pulse velocity, Windsor probe, and break-off tests. Included is a review of commonly encountered strength correlations for each of the tests. Results of compression tests of cores drilled from the two girders indicated that the c~ncrete strength after correcting for slenderness ratio effects using ASTM C42 correction factors averaged around 8 psi based upon compression tests of 4 in. diameter cores and 1, psi based upon compression tests of 2 in. diameter cores. L/d correction factors, which differed from the ASTM C42 correction factors, were derived based on test data. Of the nondestructive tests, the rebound hammer and pulse velocity tests predicted concrete strength reasonably i

4 well based upon the existing correlations. Manufacturer's relationships for the Windsor probe were not able to predict the strength. The break-off test results were found to be invalid due to the large concrete aggregate size and damage to the break-off cores during the drilling process. With one major exception, results from chloride ion tests indicated that sections of the bridge with greater exposure to deicing salts had a greater amount of chloride ion penetration. The side of a facia girder originally facing toward incoming traffic was found to have less chloride ion penetration than expected. Rain may have had a beneficial effect by washing salt residue from the facia girder during the warmer months of the year. Chloride ion penetration at the level of the reinforcing steel for the bridge girders tested was not found to be great enough to expect corrosion. ii

5 ACKNOWLEDGMENTS This research was sponsored by the Minnesota Department of Transportation and the National Science Foundation under Grant ECE and was carried out at the Civil and Mineral Engineering Structures Laboratory at the University of Minnesota. The views expressed herein are those of the writers and do not necessarily reflect the views of the sponsors. The valuable assistance of Professors Leon, Labuz and Lacabanne is greatly appreciated. Gratitude is also expressed to Sam Siegel of SDS company for the loan of the break-off equipment, Brian Pashina of Twin City Testing for providing information on concrete testing and access to equipment and personnel, and Mark Hagan at the Minnesota Department of Transportation for lending his expertise on steel corrosion in concrete and for providing chloride ion concentration data. The. writers also wish to acknowledge the efforts of the structural engineering laboratory staff, Bob Nelson, Calvin Mi tche 11 and Tom Hans en. Thanks are also expressed to Paul Bergson, Steve Olson, Kevin Klarkowski and Tim Grundhoffer for their help in the lab and with sampling and documentation. iii

6 TABLE OF CONTENTS Page ABSTRACT i ACKNOWLEDGMENTS. iii LIST -of TABLES vii LIST OF FIGURES ix Chapter 1: INTRODUCTION 1 Chapter 2: NONDESTRUCTIVE TEST METHODS Rebound Hammer 2.2 Ultrasonic Pulse Velocity Method 2.3 Combined Rebound and Ultrasonic 2.4 Windsor Probe 2.5 Break-off Test Methods Chapter 3: TEST PROCEDURE Structure Description 3.2 Test Program Test Layout Rebound Hammer Test Data Rebound Tests Before Load Testing 3 iv

7 Rebound Tests at 4 in. Core Locations Windsor Probe Test Data Pulse Velocity Test Data Break-off Test Data Four Inch Cores Two Inch Cores Core Compression Tests-2nd Beam Chapter 4: DATA ANALYSIS Concrete Compressive Strength from Core Tests Rebound Number - Strength Relationships for 2 in. Cores Corrected per ASTM C Rebound Number - Strength Relationships for Cores Corrected per Equation Pulse Velocity - Strength Relationships Windsor Probe - Strength Relationships Break-off Test - Strength Relationships Concrete Uniformity 73 Chapter 5: CORROSION OF STEEL IN CONCRETE Background Chloride Ion Concentrations in Bridge Girders Data Collection Procedure v

8 5.2.2 Sample Point Locations - Beam Chloride Ion Concentrations - Beam Sample Point Locations - Beam Chloride Ion Concentrations - Beam Sample Point Locations - Beam Chloride Ion Concentrations - Beam Base Chloride Ion Concentrations Chloride Concentrations in Bridge Deck Summary of Chloride Ion Tests 98 Chapter 6: CONCLUSIONS 1 REFERENCES 18 TABLES 114 FIGURES 141 APPENDIX A 21 APPENDIX B 24 vi

9 LIST OF TABLES Table Page Windsor Probe Compressive Strength Rebound Data (Series I, Bottom of Bottom Flange) Rebound Data (Series II, Side of Bottom Flange) Rebound Data (Series III, Side of Top Flange) Rebound Data (Series IV, Web) Rebound Data (Summary) Rebound Data (Core Locations) Windsor Probe Data Pulse Velocity Data (Before and After Coring) Break-off Test Data L/d Correction Factors 4 in. Core Strength (Corrected to Beam Mid-depth). 4 in. Core Compression Test Results 2 in. Core Compression Test Results in. Core Strength (Corrected to Beam Mid-depth) in. Cores Compression Tests 2nd Beam in. Cores Compression Tests 2nd Beam Regression Analysis (Avg Rebound # vs Strength) Rebound Number/Strength Relationships (ASTM C42 L/d Correction Factors) Rebound Number/Strength Relationships (L/d Correction by Equation 2) 21 Pulse Velocities versus 2 in. Core Strength vii

10 Table Page 22 Windsor Probe Readings vs. 2 in. Core Strength Break-off Readings versus 2 in. Core Strength Chloride Ion Concentrations (Girder 1) Chloride Ion Concentrations (Girder 2) Chloride Ion Concentrations (Girder 3) chloride Ion Concentrations (Bridge Deck) 14 viii

11 LIST OF FIGURES Figure Rebound Hammer. Rebound Number/Strength Relationships Pulse Velocity/Strength Relationships Page Combined Rebound Number and Pulse Velocity/ Strength Relationship Windsor Probe Apparatus Break-off Test Set Up Break-off Test Apparatus Break-off Value/Strength Relationship Influence of Within Test Variation and Number of Tests on Calculated Strength Bridge Superstructure Girder Gross-sectional Dimensions Prestressing Strand Location Beam Grid Layout and Test Locations Rebound Hammer Test Locations Rebound Hammer Test Patterns Rebound Hammer Test Results (Series I) Rebound Hammer Test Results (Series II) Rebound Hammer Test Results (Series III) Rebound Hammer Test Results (Series IV) Average Rebound Number (Series II, III and IV) Rebound Hammer Test Pattern for 4 in. Gore Locations 161 ix

12 Figure Page 22 Rebound Hammer Test Results at Core Locations After Fatigue Testing Comparison of Average Rebound Numbers Before and After Fatigue Testing 24 Pulse Velocity Equipment Histogram of Pulse Velocity Through Cores Before Drilling Histogram of Pulse Velocity Through Cores After Drilling. Drill Bit for Break-off Tests Drilling Apparatus Drilling Apparatus Histogram of Break-off Test Results Strength Correction Factor for L/d Ratio of 1 as a Function Measured Strength Depth-Corrected and Uncorrected Strength of 2 in. Cores Corrected 2 in. Core Strengths Versus Location Along Beam. Measured Strength vs. L/d for 2nd Beam Cores Regression Analysis, L/d vs. Measured Strength Corrected L/d Ratio Correction Factors Rebound Number/Strength Experimental Data Points, Regression Line and 9% Confidence Interval 176 X

13 Figure Page 38 Rebound Number/Strength Experimental Data Points and Regression Line Compared with Established Relationships Pulse Velocity/Strength Experimental Data Points and Population Mean Ranges Pulse Velocity/Strength Experimental Data Points Compared with Established Relationships Windsor Probe/Strength Experimental Data Points Compared with Manufacturer's Relationship 42 Comparison of NDT's Along Length of Beam Chloride Ion Concentration Sample Point Locations for Beam Chloride Ion Concentrations (Beam 1) Chloride Ion Concentration Sample Point Locations for Beam Chloride Ion Concentrations (Beam 2) Chloride Ion Concentration Sample Point Locations for Beam Chloride Ion Concentrations (Beam 3) Chloride Ion Concentration Sample Point Locations for Bridge Deck 61 Chloride Ion Concentrations (Bridge Deck) xi

14 MATERIAL PROPERTIES OF A TWENTY-YEAR-OLD PRESTRESSED BRIDGE GIRDER OBTAINED BY NONDESTRUCTIVE TESTING 1. Introduction The practice of recycling concrete structural elements which have been in service for many years requires that their current condition be evaluated. It is desirable that these evaluations be made without causing damage to the structural element. For this reason, it was of interest to investigate the ability of commonly performed nondestructive tests to predict properties of in situ concrete. Also of importance was the determination of the likelihood of corrosion of the reinforcing steel, therefore chloride ion concentrations were assessed. This study involved the use of destructive and nondestructive test methods to determine the material properties of a twenty-year-old prestressed bridge girder. The objectives of the study were: 1) to determine concrete strength of the bridge girder, 2) to assess the uniformity of concrete throughout the girder, 1

15 3) to correlate the concrete strength predicted by nondestructive test methods with the results of established destructive tests, and 4) to assess the likelihood of corrosion damage of the girder prestressing steel by determining the extent of chloride ion penetration into the concrete. Chapter 2 describes the nondestructive tests and strength correlations available in the literature. Chapter 3 presents the experimental program and results of the nondestructive and destructive tests. The relationship between nondestructive and destructive test data is discussed in Chapter 4. Chapter 5 presents the information related to corrosion of reinforcing steel as well as the chloride penetration data and Chapter 6 presents conclusions. 2

16 2. Nondestructive Test Methods Nondestructive test methods have been developed to assess in situ concrete quality. Such tests include surface hardness techniques, penetration resistance methods, pullout tests, break-off tests, ultrasonic methods, and combined methods. These techniques do not measure concrete strength directly; however, each method provides information about concrete quality which may then be related to concrete strength using a calibration procedure. For existing structures, the procedure typically involves drilling an appropriate number of concrete cores from the structure and performing core compress ion tests on the cylinders. This data and the results from the nonde s true ti ve tests in the vicinity of the core can then be used to prepare a calibration curve for the entire structure. ~our nondestructive techniques were used in the present study to evaluate which of them provided the best estimates of concrete strength. These methods, rebound hammer (surface hardness), Windsor probe (penetration resistance), ultrasonic pulse velocity, and break-off techniques are described in detail in the following sections. Relationships between nondestructive test results and concrete compressive strength developed by other investigators are also presented. 3

17 2.1. Rebound Hammer The rebound technique is a quick, easy test that is generally used to assess the uniformity of concrete. It does not directly measure concrete strength, but rather indicates the surface hardness of concrete by measuring the rebound of a mass impacting the concrete. There is little theoretical relationship between concrete strength and the rebound measurement because the rebound number is a function of surface crushing and internal friction of the material. However, attempts have been made to correlate rebound test results with concrete strength alone and in conjunction with other nondestructive tests. The rebound hammer was developed in 1948 by a Swiss engineer, Ernst Schmidt, thus the sometimes used name "Schmidt Hammer". The hammer (Fig. 1) utilizes a spring loaded mass that is released when the hammer is slowly pressed against the concrete surface. The mass tmpacts the concrete with a specified amount of energy and the rebound distance of the mass (the rebound number) is given on a sliding scale. The test can be performed on surfaces oriented at any angle. At each angle, a different rebound number would be observed, and thus different calibration curves must be used. In early research with the rebound hammer, Greene [1] found the within test variation to be 1.5 percent for 12 specimens at age 7 days and 7.4 percent for 18 specimens at age 28 days. The within test 4

18 variation is defined as the coefficient of variation within a sample of tests (See Appendix A Statistical Calcula- tions). Malhotra [2] lists some of the variables that cause rebound hammer readings to differ for concretes of the same strength. These include: -Smoothness of the concrete surface. More accurate results can be obtained by grinding rough surfaces to a uniform smoothness. -Type of form. Troweled surfaces or surfaces made agains.t a metal form will give rebound readings 5-25 percent higher than surfaces made against wooden forms because of differences in surface smoothness. -Size, shape and rigidity of the specimen. The s trueture should be supported to restrict elastic deforma.tions. -Moisture condition of the concrete. Wet concrete gives readings approximately 5 percent lower than dry concrete. -Type of coarse aggregate. Concretes made with crushed limestone give rebound numbers approximately 7 points lower than cone rete s made with grave 1 coarse aggregate. 5

19 -Type of cement. Rebound numbers will be significantly different between concrete made with ordinary portland cement and concretes made with other cements (highalumina cement, supersulfated cement). -Carbonation of the concrete surface. Carbonation effects are most severe in older concretes where the layer of carbonated concrete can be several millimeters thick. Rebound numbers can be significantly higher in such concretes. Rebound readings are affected by surface anomalies such as a large piece of aggregate close to the surface, a near surface void or reinforcing steel near the concrete surface. For this reason, a number of rebound readings, usually about nine or ten, should be taken in a particular area and averaged after throwing out abnormally high or low readings to obtain a single rebound value for the area [3]. Although the m<;~.nufacturer of the test hammer provides an empirical correlation between rebound number and concrete strength, most authors caution against using these values. Zoldners [4] recommends calibrating rebound hammer readings with compression tests of concrete cylinders for each type of concrete evaluated. His procedure was to obtain a representative number of readings on the sides of a 6 in. by 12 in. test cylinder that was supported in a compression machine at a load of 15 psi prior to breaking the cylinder. The 6

20 average of the rebound readings, disregarding the highest and lowest readings, was then plotted against cylinder test strength. For his test program of 5 standard cylinders, Zoldners found that rebound readings predicted cylinder strength within ± 15 percent. The general consensus among authors was that for properly calibrated tests under cont~olled conditions, the accuracy of the rebound test ranged from± 15 to 2 percent. For rebound readings taken on a horizontal surface the relationship given in the manufacturer's curves is approximately described by the linear relationship: f' c 24.6 R for 2 < R < 65 (Eqn. 1) where f'c is the estimated concrete strength in psi and R is the average rebound reading for a test area. Samarin and Dhir [5] also propose a linear relationship between rebound number and strength where the constants are obtained by a calibration procedure for each specific type of concrete. A procedure reported by Facaoaru [6] is widely used in Romania to determine in situ cone rete strength and is summarized be low. A rebound number/strength relationship for a previously defined "standard concrete" characterized by cement type, aggregate type, maturity and moisture content is given 7

21 by: f'c ref R 2 2 (f'c ref in psi) (Eqn. 2) Then, if the concrete tested is not the reference concrete:. f' c f'c ref x Ct (Eqn. 3) The constant Ct is obtained by one of three methods: 1) experimentally by testing cores (Ct = f' c experimental/f' c ref), 2) empirically if the concrete composition is known by multiplying a number of coefficients based on the concrete composition (type and amount of cement, nature of aggregate, maturity and moisture condition) to obtain Ct, or by 3) a combined method where Ct is given by 2/3 Ct experimental plus 1/3 Ct empirical. Facaoaru estimates the accuracy (range in which 9 percent of the results will occur) of the three.techniques as follows: 15-2% experimental method 18-28% empirical method 12-18% combined method Facaoaru and Samarin and Dhir propose a more accurate method using the rebound hammer method in conjunction with pulse velocity testing. The combined nondestructive method 8

22 and its accuracy is discussed in Section 2.3. Figure 2 gives a graphical comparison of the rebound number/strength relationship given by the manufacturer (Equation 1) and the relationship given by Facaoaru (Equation 2). The curve depicting Facaoaru's relationship is given for the "standard concrete" as defined by this method Ultrasonic Pulse Velocity Method The pulse velocity method is based upon measuring the time of transit of an electronically generated longitudinal (or compression) wave through concrete. Time of transit is divided into the path length to obtain the wave velocity for concrete. The main components of the pulse velocity equipment are the timing circuit and the sending and receiving piezoelectric transducers. The transducers convert electrical into mechanical energy and vice versa. Frequencies of the vibrations are determined by selecting the characteristics of the transducers. Path lengths may range from two inches to fifty feet. Frequencies used are in the range of ten to fifty kilohertz or greater. The lower frequencies are used for the longer path lengths to minimize attenuation. Higher frequencies are used to obtain greater accuracy for shorter path lengths [7]. 9

23 The pulse velocity/strength relationship given in the literature is generally either an exponential function or a power function [6, 8, 9, 1]. Theoretically, the compression wave velocity in a bar is given by the expression: v J E/p (Eqn. 4) where, E modulus of elasticity of the material V p wave velocity density of the material In an infinite medium, where lateral displacement is restrained, the expression is: v J E/Kp (Eqn. 5) where, K is a constant of the material derived from its Poiss~n's ratio, v: K (l+v)(l-2v)/(l-v) (Eqn. 6) Poisson's ratio for high strength concrete, as well as for normal stren~th concrete, has been reported to be close to.2 [11]. Whitehurst [8] suggests that the dynamic Poisson's ratio be assumed as. 24 when determining material properties by dynamic methods such as pulse velocity. The 1

24 constant K is.85 for Poisson's ratio of.24 and K is.9 for Poisson's ratio of. 2. It can be seen from the above relationships that pulse velocity is independent of wave frequency. However, there is controversy regarding whether K should vary depending on the size and shape of the specimen. Some authors feel that three different values should be used: K=l for bars, K=l-v 2 for slabs and K=(l+v)(l-2v)/(l-v) for mass concrete. Others hold that measured velocity is essentially independent of variations in size, shape and path length and that the latter value should be used in all cases [8]. ASTM Standard C597 for pulse velocity in concrete simply defines K as a constant and does not provide any further guidance [7]. Because concrete compressive stre,ngth has been shown to be related empirically to its secant modulus of elasticity, E, it is possible to relate pulse velocity to compressive strength. Pauw [12] gives the empirical expression for E that is used in the ACI code: E (Eqn. 7) where E and f' c are both in psi and w is the weight of concrete in pounds per cubic foot (pcf). Taking was 145 pcf 11

25 for normal weight concrete leads to: E 576 J f'c (Eqn. 8) Combining Equations 5 and 8, using w = 145 pcf, and converting units, the following is obtained: f' c 2.94E-13 K 2 V4 (Eqn. 9) with V in units of ft/s. Gardiner and Hatcher at Washington University [9] tested 18 cylinders to correlate pulse velocity to compressive core strength using an equation similar to Equation 9 above and found K to be an average of approximately. 7. This was lower than the K-value of. 85 that was obtained when K was calculated using the equation for mass concrete and Poisson's ratio of.24. A Poisson's ratio of.32 is implied if K is taken as.7. This value is somewhat higher than is generally assumed. Upon performing a regression analysis of the equation, they obtained a correlation coefficient, r, of. 72 and r 2 was. 52. This indicated a fairly good correlation between pulse velocity and compressive strength (See Appendix A Statistical Calculations). Other investigators have proposed different relationships between pulse velocity and compressive strength. Facaoaru, 12

26 defines a standard reference concrete as he does for his rebound hammer correlations discussed in the previous section [ 6 ]. He models the relationship between pulse velocity and compressive strength with the exponential relationship given below:. f' c a ec3.35xloe-4v) (Eqn. 1) where the coefficient, a, is given by: a (Eqn. 11) and where aref for the refere~ce concrete is defined as 34.8 psi and V is in the units of ft/s. Ct is determined by one of three methods similar to the methods used to determine Ct for the rebound hammer/strength relationship given by Facaoaru. Ct can be determined 1) experimentally by testing cores,. 2) empirically if the concrete composition (type of cement, type and size of aggregate, amount of cement, characteristics of fine aggregate, moisture condition, maturity, and admixtures) is known or, 3) by a combination of the two methods using the weighted mean average of Ct experimental and Ct empirical. According to Facaoaru the accuracy (range in which 9 percent of the results will 13

27 occur) of the method is: 14-18% experimental method 18-25% empirical method 12-16% combined experimental/empirical method. The pulse velocity relationships discussed above are shown in Figure 3. Pulse velocity/strength relationships may be established for concrete but they are influenced by factors such as concrete composition, age, moisture condition, curing conditions, cracking, and deterioration. Sturrup, Vecchio, and Caratin summarized how different variables affect strength and pulse velocity [ 13]. They found that the relationship varied with properties of the concrete components, concrete mix proportions, curing history, moisture condi t.ion, strength: and deterioration. In general, at the same 1) Pulse velocity measured through lightweight aggregate concrete was much lower than that measured through normal weight aggregate concrete. For normal weight concrete, pulse velocity was also affected by aggregate type (gravel, crushed limestone) but the 14

28 effect was not as significant as that between normal and lightweight concretes. 2) Pulse velocity measured through high aggregate content concrete was greater than that measured through a lower aggregate content concrete. 3) Pulse velocity measured through concrete with large aggregate was greater than that measured through concrete with a smaller-sized aggregate. 4) Pulse velocity increased as water/cement ratio decreased. 5) Pulse velocity increased with moisture content. Additionally, they stated that when subjected to freeze-thaw cycles, concrete exhibits a greater decrease in pulse velocity than in strength [13]. This is probably due to microcracking which may affect strength and pulse velocity differently. Tests with three instruments and five operators indic~ted that cracking also affects the variability of test results [7]. Through sound concrete, the same operator with different equipment or different operators with the same equipment achieved good repeatability such that differences in test results were within 2 percent; whereas, through cracked concrete, differences between test results increased to as high as 2 percent. The presence of steel or voids in the concrete also affects pulse velocity because the wave 15

29 velocity through steel or air is different from the wave velocity through concrete. Due to the variables affecting pulse velocity in different concretes, calibration is required in order to obtain a satisfactory pulse velocity/strength relationship. ASTM Standard C597 [7] recommends that pulse velocity be used for assessing uniformity of concrete, determining the presence of voids and cracks, estimating the severity of deterioration but not for measuring concrete strength or modulus of elasticity unless calibration is conducted to determine the pulse velocity/strength relationship of the concrete in question Combined Rebound and Ultrasonic Methods ro obtain better correlations than those possible with only one nondestructive test, investigators have used multiple regression equations with rebound hammer and pulse velocity data as the independent variables and compressive strength as the dependent variable. These models developed by Facaoaru, Tanigawa et al. Samarin and Dhir and others working independently were more accurate than strength relationships for the individual nondestructive tests [5, 6, 1, 14]. 16

30 Facaoaru developed a series of iso-strength curves plotted on a horizontal pulse velocity axis and a vertical rebound number axis for a standard reference concrete [ 6]. The strength obtained from these curves is multiplied by a factor (Ct) to correct for nonreference concrete. As is the procedure for both rebound hammer and pulse velocity, Ct can be obtained either experimentally, empirically or by a combination of the two methods. The accuracies (range in which 9 percent of the results will occur) for the combined rebound hammer and pulse velocity method are: 12-16% experimental method 15-2% empirical method 1-14% combined experimental/empirical method An example of Facaoaru's curves is shown in Figure 4. Regression analysis equations found in the literature take a number of forms such as: f' c AR + BV + C (Eqn. 12) where A,B and C are constants, R is the rebound number and V 17

31 is the pulse velocity. Another form of the regression equation used by some investigators is: ln f'c AR + BV + C (Eqn. 13) Tanigawa et al. [1] tested 82 specimens by pulse velocity and rebound methods. He then determined the concrete compressive strength using a hydraulic testing machine. The following correlation coefficients, r, for multiple regressions on Equations 12 and 13, above were obtained: Eqn. 12, AfO, B=O r =.784 Eqn. 12, A=O, BfO r =.545 Eqn. 12, AfO, BfO r =.936 Eqn. 13, AfO, B=O r =.788 Eqn. 13 AfO, BfO r =.938 These. values of r indicate good correlations (see Appendix A) except for Equation 12 with A equal to zero. The low correlation coefficient for Equation 12 with A equal to zero was not surprising because the pulse velocity/strength relationship is generally taken to be either a power or an exponential function. The correlation using the two variables was much better than using any of the variables alone. 18

32 Samarin and Dhir took the general form of the regression equation to be: f' c AR + BV 4 + C (Eqn. 14) and found the 95 percent confidence interval ranges for conc~ete compressive strength to be within 3 psi [5] Windsor Probe The Windsor probe, like the rebound hammer, is a hardness tester. The method consists of firing a powder activated probe with a known amount of energy into the concrete and measuring the exposed probe length. The Windsor probe apparatus includes the probe, the powder charge, the gun or driver, a firing template and a measuring device (Fig.. 5). There are,two types of probes available: a "gold" probe (3-1/8 in. length by 5/16 in. diameter) for lightweight concrete and a "silver" probe (3-1/8 in. length by 5/16 in. diameter with the penetrating end diameter reduced to 1/4 in. for 9/16 in. of the length) for normal weight concrete. The manufacturer also provides a method to adjust the power level by making an adjustment in the driver. Standard power is for mature concrete with a compressive strength greater than 32 psi. Low power is used if the 19

33 compressive strength is less than 32 psi. Three probes in a given test area are averaged to obtain a single data point. ASTM Standard C83 for the Windsor probe test states that probes should not be fired within 4 in. of the edge of the concrete surface nor within 7 in. of each other [15]. The maximum difference among the three individual probe measurements allowed by the ASTM C83 precision statement for the Windsor probe test is.33 in. for concrete with maximum 1 in. aggregate size. If any two separate sets of three measurements are compared, a difference of.16 in. or greater between the two values indicates a significant difference in the concrete in the two areas [15] The manufacturer recommends a method to calibrate the test results by determining the hardness of the aggregate. The Mohs' value of the aggregate is obtained using a Mohs' test kit consisting of nine numbered minerals which make up a scale of hardness. The aggregate is sera tched with the hardest mineral, then with each successively softer mineral until the stone is found that will not scratch the aggregate. The number on the last stone is the Mohs' hardness of the aggregate. A table is provided giving concrete compressive strength for the exposed probe length and the Mohs' number of the aggregate (Table 1). One problem with this method is that a particular concrete may contain aggregate with a range of Mohs' hardness, thus making it difficult to 2

34 pinpoint the concrete compressive strength. Malhotra, in reviewing the literature and in his own investigations, found that the manufacturer's tables were not satisfactory as they sometimes greatly overestimated and sometimes greatly underestimated the actual strength [2] ASTM Standard C83 for the Windsor probe states that the test is best used for assessing uniformity of concrete, delineating zones of poor quality and indicating strength development as supplementary information but not as the only measure of concrete strength [ 15]. Relationships between concrete compressive strength and Windsor probe data may be established by strength tests of cores taken from a structure. However, test cylinders that have been probed should not be used to determine compressive strength due to the damage caused by probing [2]. The within test variation for the Windsor probe has been ~eported to be from percent [2] Break-off Test The break-off test method was developed and patented in Norway in The test involves applying a cantilever bending moment to a 2.13 in. diameter by 2.75 in. long cylinder that has been either cast or drilled into the 21

35 concrete. A force is applied at the top of the cylinder parallel to the concrete surface and the pressure required to break the cylinder off below the concrete surface is measured. Figure 6 illustrates this set up. The testing apparatus consists of a load cell connected to a handoperated hydraulic pump and a manometer which measures h y d r a u l i c p r e s sure ( F i g. 7 ). In o r de r to fa c i l i tate t e s t in g in both high and low strength concretes, one of two possible test ranges may be selected. Low level ranges from -36 psi and high level from -19 psi. The manufacturer provides a graph giving an empirically derived relationship between break-off manometer pressure and concrete compressive strength [16]. The relationship for the high level is pictured in Figure 8. Regression analysis performed on this curve gave the equation:.f, c 1.4 B 1 76, (Eqn. 15) where BO is the break-off reading. The correlation coefficient, r was Because the curve is a power function, this equation seems to indicate that the test becomes less sensitive at higher concrete strengths. The manufacturer recommends taking the average of five break-off readings from a single structural member to obtain 22

36 an average break-off value for the member. The within test variation for five break-off tests is approximately 9 percent. Based on this, the characteristic concrete strength can be predicted on approximately an 85 to 75 percent confidence level depending on the coefficient of variation of the concrete strength in situ (see Figure 9). For core compression tests where the coefficient of variation for three tests is about 3 percent, the corresponding prediction would be between approximately 95 and 6 percent. The accuracy of the break-off test method is less than for core tests. However, this can be compensated for by using the greater number of break-off tests (5 opposed to 3). For a material with a coefficient of variation in situ greater than 1 percent, five break-off tests should give a better estimate of concrete compressive strength than three cylinder compression tests [17] as shown in the same figure (Fig. 9). In new concrete the cylinders can be produced by using removable plastic forms. However, for existing concrete the cores must be drilled. Most available data consists of a comparison of break-off strengths for formed break-off cylinders to standard compression cylinders poured at the same time. The correlation in an older existing structure between drilled break-off cylinder values and drilled compression cores has not been adequately investigated. It 23

37 has been reported that the variability of the test increases slightly for drilled cylinders over cast cylinders but that the correlations between the break-off test and compressive strength were not affected [18]. Another factor found to affect variability is the maximum size of the coarse aggregate. Barker and Ramirez [ 18 ] compared break-off strength for formed break-off cylinders with standard cylinder compressive strength for different concrete mixes at 3, 7, 14, 21, and 28 days and found that: 1) 1 in. aggregate did not affect the regression equation but greatly increased the uncertainty (confidence intervals) over that observed for 1/2 in. aggregate, 2) water-cement ratio did not affect the break-off correlation over the range investigated, 3) the type of aggregate did not affect the correlation and 4) their results were of the same magnitude as the manufacturer's correlation curve. 24

38 3. Test Procedure 3.1. Structure Description This test program was carried out on a prestressed bridge girder from a center span of the two-lane Boone Avenue overpass that crossed Interstate 94 in Brooklyn Park, a suburb of Minneapolis, Minnesota. The bridge, constructed in 1967, was taken down in order to widen the road. After a short period of time during which the bridge girders were stored outdoors at a construction yard, four of the bridge girders were moved into the large structures laboratory at the University of Minnesota in February, The bridge was designed for HS2 loading in accordance with the 1961 AASHO design specifications. The bridge superstructure consisted of two center 65 ft. spans of five girder.s each and two outer 42 ft. spans of three girders each (Fig. 1). The beam used for the nondestructive tests was an interior girder from one of the 65 ft. center spans. In order to provide additional data, cores were later obtained from a second identical girder from the bridge. The core compression test results from the second beam are discussed in section immediately following the two sections giving core compression test results from the first beam. 25

39 Girder cross-sectional dimensions are given in Figure 11. The nominal 65 ft. long spans were actually 64 ft. 8 in. out-to-out of the girder. Girders were reinforced with thirty one-half inch diameter, seven wire prestressing strands, of which twenty-two straight strands ran through the bottom flange. The remaining eight strands were draped from the top flange to hold down points that were located approximately five feet from the centerline of the girder (Fig. 12). Stirrups (No. 4 bars) were located on approximately 1 in. centers along the length of the beam. Initial prestress for each girder was specified as 756 kips. Twenty-eight day concrete compressive strength averaged 67 psi for all of the bridge girders. Specified concrete strength at the time of transfer was 45 psi. Concrete mix data is given below: Cement type: Type III high early strength Coarse Aggregate: 1 in. max. size river gravel Fine Aggregate: Fineness Modulus 2.54 Water/cement ratio:.31 26

40 3.2. Test Program Test Layout Nondestructive tests were performed at each of twelve locations on the web along the beam. The locations were chosen to avoid stirrups and draped prestressing strands as well as cracks formed previously during fatigue testing. For each area, a series of tests were run, compared with other tests, and compared with compression tests on cores taken at the same locations. Each location was numbered from one through twelve in order to reference the location and all of the tests performed at that location. For example, if three Windsor probes were shot at location 5, they would be labelled WPS-1, WPS-2, and WPS-3. In order to easily record test locations, a grid was set up with vertical grid lines along the length of the beam. The grid lines were spaced at 2 ft. intervals except for the end grid lines which were placed at 28 in. from the ends of the beam (Fig. 13). Figure 13 also shows the locations of all the nondestructive tests. Grid lines were numbered from one to thirty-one and grid blocks were referenced by their bordering grid lines. The two sides of the beam were labelled A and B. 27

41 With the exception of a group of rebound hammer tests, the series of nondestructive tests were run after the beam had been subjected to fatigue cycles and then statically loaded to failure in the laboratory. A portion of the center section of the beam was unsuitable for testing due to extreme cracking which resulted from the load tests. There were a limited number of uncracked regions in the web which did not contain reinforcement and were of sufficient dimensions to allow the destructive and nondestructive tests. Consequently, of the twelve locations selected for testing, half were located in cracked concrete and half in uncracked concrete. In the cracked concrete, tests were laid out such that cracks were avoided. In order to select locations such that reinforcing steel would be avoided, a pachometer was used to locate both the vertical stirrups and the draped prestressing strands in the web. Tests were planned such that for each test location either one or two 4 in. compression cores were drilled depending upon spacing constraints. Five break-off tests were performed and three Windsor probes were shot as the layout permitted. In addition, rebound hammer readings and pulse velocity readings were taken at each core location. 28

42 Appendix B contains diagrams of each grid block showing all reinforcing steel, core locations and test locations for the beam tested Rebound Hammer Test Data Two sets of rebound tests were performed. The first was completed on the beam flanges and web prior to load testing of the beam. The second set was taken at core locations after beam failure. A Schmidt N-34 hammer was used and ASTM Standard C85 for the rebound number of hardened concrete was followed [3]. For all tests a single rebound value was obtained by averaging a number of separate rebound readings taken in the test area. Any test readings that varied more than seven rebound points from the average value were discarded and the remaining values re-averaged. If the test surfac.e appeared crushed or broken, this was noted and the test value was discarded. 29

43 Rebound Tests Before Load Testing This set of rebound tests consisted of four series of rebound values along the length of the beam: Series I - bottom surface of the bottom flange Series II - Series III - Series IV - side of the bottom flange side of the top flange web Nine separate readings were averaged to obtain the rebound value for each test area. The tests patterns (Figs. 14 and 15) were centered in each grid block and test pattern locations were referenced by the corresponding grid lines. Tables 2-5 give the average, minimum and maximum rebound value for each set of nine readings taken along the beam. Values shown in the tables were obtained after any bad readin.gs were discarded using the reject criteria. Twentythree out of 558 (approximately 4 percent of) readings for test series I-IV were discarded. Rebound readings for series I cannot be directly compared with the other rebound readings because they were taken in a vertically upward direction and the others were taken horizontally. Results are depicted graphically along the length of the beam in Figures 16 through 2. Table 6 summarizes results for test series I-IV and gives the average, absolute maximum and 3

44 absolute minimum rebound values for each test series. A rebound value was the average of the nine individual readings for each test pattern. Table 6 shows that test series III and IV for the top flange and the web exhibited greater variability (COV's of 3.56 and 3.14 percent respectively) than series I and II for the bottom flange (COV's of 1.3 and 1.41). Additionally, the average rebound value for series III and IV was less than for series II. This may be due to water gain effects causing a variation of concrete strength in a vertical direction with weaker concrete toward the top and stronger concrete toward the bottom. The average rebound number for series IV in the web was Based on the manufacturer's equation (Equation 1), this value corresponds to a concrete strength of 798 psi. The coefficient of variation for the concrete strength was 4.35 percent Rebound Tests at 4 in. Core Locations Rebound values after fatigue testing were obtained for each drilled core by taking nine readings from each side of the core face (a total of eighteen rebound points) prior to drilling. Rebound points were taken in a square pattern spaced two inches apart such that the center of the square coincided with the center of the core which was subsequently 31

45 drilled (Fig. 21). Table 7 shows results obtained for each core location. Average, maximum and minimum values are shown for each set of eighteen readings after discarding outlying values according to the reject criteria. Eight out of 347 (2.3 percent of) readings were discarded. Average results along the beam are also shown. These compare almost identically with results obtained along the beam prior to fatigue testing. The average rebound number obtained along the web was 54.1 prior to fatigue testing (test series IV) compared with 54. after testing. These average rebound numbers correspond to concrete strengths of 798 and 796 psi based on the manufacturer's curve (Equation 1). Coefficients of variation for rebound numbers along the beam were also relatively close: 3.14 percent before testing compared with 3. 2 percent after testing. Coefficients of variation for the concrete strengths based upon the manufacturer's equation were 4.35 percent prior to fatigue testing and 4.17 percen~ after fatigue testing. Figure 22 shows average, minimum and maximum rebound numbers along the beam at core locations after fatigue testing. Figure 23 compares average rebound numbers before and after fatigue testing. 32

46 Windsor Probe Test Data Windsor probe testing was conducted in accordance with ASTM standard C83 [15] The Windsor probe test system was used on standard power with "silver" PRS-1 probes. Three probes were fired within six inches of the core locations for each area using the single probe template provided with the equipment. A three probe template with a mechanical averaging device is generally used; however due to space constraints, the single probe template was used for this work and exposed probe lengths were measured individually. The testing procedure was as follows. Three points near the planned core locations were identified for each area. For spall control, a circle was scored into the concrete surrounding the point. The probes were then fired into the concrete. To measure the exposed probe length, any raised porti~ns of the concrete surface around the probe were chipped off. A measuring base plate was put on the concrete surface by inserting the probe through a hole in the base plate. A measuring cap was screwed on the probe and four measurements from the cap to the base plate were made around and parallel to the probe. The four measurements were averaged to obtain the exposed probe length. This procedure was repeated for the two other probes in each test area. If the maximum range of the readings for each area did not 33

47 exceed.33 in. the three values were averaged to obtain a single Windsor probe reading. If the maximum range was greater than. 33 in. the two values falling within the required range were averaged. The. 3 3 in. index is the maximum allowable range for groups of three individual measurements given by the precision statement in ASTM C83 [ 15] : Windsor probe results are shown in Table 8. A probe was not shot for three of the test areas due to space constraints. In two cases, the three individual Windsor probe measurements did not fall within the required range of.33 in. This resulted in only two measurements being used to obtain the average value in three cases and one case in which an average value could not be obtained. The average exposed probe length for nine values obtained from the average of three measurements and three values obtained from the ayerage of two measurements was 2.19 in. with a minimum of 2.5 in. and a maximum of 2.31 in. The coefficient of variation for the eleven values along the web was 3.6 percent. The Mohs' hardness of the aggregate ranged from 6 to 7. The tables provided by the manufacturer correlating exposed probe length to compressive strength are described 34

48 by the following linear relationships: f' c 8 WP - 16, for Mohs' No. of 6 (Eqn. 16) 'f, c 8749 WP , for Mohs' No. of 7 (Eqn. 17) wher~ f'c is in psi and WP is the exposed probe length in inches. Based on these relationships, the exposed probe lengths corresponded to an average compressive strength of psi, a minimum strength of psi and a maximum strength of psi. The coefficient of variation for compressive strength along the web based on these equations was percent. These values, which were obtained by directly applying the manufacturer's tables, are compared with core compressive strength in Chapter Pulse Velocity Test Data A James instruments F-meter (Fig. 24) was used to obtain pulse velocity readings at the center of each 4 in. core both prior and subsequent to drilling the core. Transducers with vibrational frequencies of 5 kilohertz were used. Transit times were measured to the nearest tenth of a microsecond. After the cores were drilled, path lengths 35

49 were measured to the nearest thousandth of an inch. Path lengths were divided by transit times to obtain pulse velocities in ft/s. Based upon the accuracy of the time and distance measurements, the sensitivity of pulse velocities was calculated and found to be within ± 2 ft/s for the range of time and distance measurements recorded. Table 9 shows pulse velocity values obtained before and after drilling. Figures 25 and 26 are histograms of pulse velocity through the cores before and after drilling. The mean pulse velocity before drilling was 15, ft/s, the standard deviation was 199 ft/s and the coefficient of variation was percent. The mean pulse velocity after drilling was 15,1 ft/s, the standard deviation was 29 ft/s and the coefficient of variation was 1.38 percent. If a pulse velocity/strength relationship is assumed, then 'l;:he value and accuracy of f' c can be calculated from the pulse velocity measurements. However, the theoretical relationship (Equation 9) and Facaoaru' s relationship (Equation 1) give greatly different results when applied directly without considering additional data such as core compressive strength or concrete composition. Equation 9 gives f' c of 1,75 psi for a 15, ft/s pulse velocity assuming a K-value of.85 and f'c of 12,6 psi assuming a K-value of.9. The resulting strength for K.85 is 36

50 accurate to within +589 psi or -559 psi based upon the sensitivity of the measurements. The resulting strength for K =.9 is accurate to within +656 psi or -63 psi based upon the s ens i ti vi ty of the measurements. If the cone rete under investigation is assumed to be the same as the reference concrete as defined by Facaoaru, then Equation 1 give~ an f'c value of 5318 psi for 15, ft/s pulse velocity with an accuracy of +369 or -345 psi based upon the sensitivity of the measurements. Pulse velocities before and after drilling were compared in order to determine if any damage was caused to the 4 in. cores by the drilling process that would result in core strengths not representing in situ strength accurately (Table 9). The ratio of pulse velocity after drilling to pulse velocity prior to drilling ranged from. 99 to 1. 3 with an average of The 1 ft/s difference between average pulse velocities before and after drilling was not significant based upon the accuracy of the measurements. No damage was caused to the cylinders by the drilling process that could be detected considering the measurement sensitivity. 37

51 Break-off Test Data Break- off cyl in de r s web of the girder using a were drilled horizon tally in the diamond drill bit available from S cancem, the manufacturer of the break- off equipment. The bit was specially designed with two drilling surfaces to drill- the counter-bored cylinder required for the break-off tests (Fig. 27). A Milwaukee Dyma-Drill with a vacuum base was used to drill the cylinders. The drill had the following specifications: 75 maximum rpm, 12 Watts, 4 in. maximum allowable core diameter, water cooled. The drill was mounted on an automobile jack that was in turn mounted on a small cart with wheels to facilitate positioning of the drill (Figs. 28 and 29). In order to drill a cylinder at a selected location, the drill was positioned such that the bit was centered at the appropriate location, the vacuum pump was turned on and the base was pressed against the concr~te surface to allow the vacuum to support the drill. Cooling water was turned on and drilling of the core commenced. Drilling time for the majority of the break-off cylinders was about 5-1 minutes. However, the drill times for a few of the cylinders were longer (2-3 minutes) with two or three cylinders taking as long as 45 minutes to drill. The longer drilling times were caused by not being able to obtain an adequate suction base seal on rough or cracked sections of the web. This resulted in the inability 38

52 to transfer sufficient force to the drill bit without tearing the drill base from the concrete surface. This problem was partially solved using a cable to provide additional support. In cases where the concrete was extremely cracked anchor bolts were drilled to hold the base to the concrete. The manufacturer gives a drilling time of 2-1/2 minutes for a cylinder of compressive strength 3625 psi at a drill speed of 16 rpm. Because the drill used in this investigation was operated at a lower rpm and the concrete had a greater strength, the usual drilling time of 5 to 1 minutes seemed reasonable. Locations for the break-off tests were planned such that five break-off tests would be located in the vicinity of each set of cores. Reinforcing steel, spalled concrete caused by Windsor probe testing, and cracks had to be avoided when positioning the drill base. In some cases, the cylinqer broke off prematurely due to the loss of base suction during drilling. Because of these problems all five cylinders in some sets could not be drilled. A total of 55 break-off cylinders were drilled: nine groups of five cylinders, two groups of four cylinders and one group of two cylinders. The latter group was in an extremely cracked section of the beam. The layout for the break-off tests is shown in Appendix B. 39

53 The Scancem break-off tester was used to obtain breakoff values after all the cylinders had been drilled. The procedure recommended by the manufacturer was followed to carry out the tests. All tests were performed at the high test level (the test scale ranged from to 19 psi). The load cell was placed into the drilled slot and after closing the pressure release valve and resetting the gage needle to zero, hydraulic pressure was applied by pumping one stroke per second. Each stroke raises the pressure approximately 5 bars. Pumping was continued until the cylinder broke at which time the pressure gage reading was recorded. The tester calibration was checked intermittently using the proving ring provided with the equipment. Table 1 gives the break-off tester gage readings for each cylinder along with the average value and standard deviation for each group of break-off readings. Also given in Table 1 is the compressive strength corresponding to the average reading for each group of tests based upon the manufacturer's break-off reading/strength correlation. The values of f'c calculated were significantly lower than core compressive strengths measured (core test results are discussed in Sections and ) and were even less than the twenty-eight day concrete compressive strength originally reported for the girder. The manufacturer's correlation was based upon formed cylinders and therefore 4

54 may not account for any damage to the core from drilling. This could be a factor in the break-off values being lower than expected. Furthermore, data from prior research shows that larger aggregate size increases the variability of break-off results [18] and this may be an additional factor affecting the results. Figure 3 is a histogram of all break-off test results Four Inch Cores Twenty separate nominal 4 in. diameter cores were drilled from the web of the girder on which the nondestructive tests had been performed. Cores were drilled using a 4 in. diamond drill bit and the same drilling equipment as used to drill the break-off cylinders. Drilling time for each cylinder was approximately 2 minutes except when the drill was poorly supported as explained in the previous section, in which case, drilling times were significantly higher. With the exception of four locations, two cores were drilled at each test area. Single cores were drilled at locations 3, 7, and 8 due to spacing constraints. Although two cores were planned for location 12, only one was drilled because of the presence of steel reinforcement. Cores were capped with a sulfur cement; core lengths after capping and core diameters were measured and recorded. 41

55 Because the web thickness was approximately 7 in. it was not possible to drill standard 6 in. diameter by 12 in. long cores through the web. Therefore, a 4 in. diameter core drill bit was used to achieve a length/diameter (L/d) ratio as close as possible to the standard L/d ratio of 2. Core lengths after capping averaged 7.5 in. with a minimum of 7.4 in. and a maximum of 7.7 in. Core diameters averaged 3.82 in. L/d ratios averaged 1.96 and ranged from 1.93 to 2.1. It was necessary to relate the compressive strengths of the nonstandard 4 in. cores from the web of the bridge girder to the strengths of standard 6 in. by 12 in. cores. Two effects were examined: 1) the effect of a nonstandard diameter (approximately 4 in. versus a standard of 6 in.) and 2) the effect of a nonstandard L/d ratio ( versus a standard of 2.). Gardiner and Hatcher [ 9], conducting a series of nondestructive tests similar to those in this investigation, cast and tested a series of nonstandard and standard cylinders of different compressive strengths and found the ratio of compressive strengths of nonstandard 4 in. by 8 in. cylinders to standard cylinders to average This average was observed over a range of compressive strengths from 555 to 833 psi. They therefore concluded that no correction factor was necessary to account for differences 42

56 in diameter given the same L/d ratio for the particular size cylinders tested. This result is supported by other researchers who found that this was true as long as the core diameter was at least three times the maximum coarse aggregate size (19]. The maximum aggregate size for cores in the study reported here in was 1 in., therefore a core diameter of 3.8 in. was satisfactory and no correction for core diameter was applied. As L/d ratio decreases from a value of 2, measured core strength increases. In general, the measured core strength is expressed as an equivalent strength at an L/d ratio of 2 by multiplying the measured strength by a correction factor. ASTM Standard C42 limits core testing to L/d ratios between 1 and 2 and lists the correction factors to use when L/d ratios are less than 1.94 (no correction is required by ASTM C42 when L/d is between and 2. ) [ 2] The ASTM C42 correction factors are applicable for nominal strength concretes from 2 to 6 psi and are shown in Table 11 along with the British Standards Institution correction factors. ASTM C42 states that correction factors depend on strength and that the recommended correction factors represent an average over the range of strengths [ 2]. Research indicates that less correction is required as concrete strength increases due to the fact that the magnitude of the restraint developed between the core and 43

57 the testing machine platen is a function of the relative stiffness of the two materials and concrete stiffness is in turn a function of strength [19]. The strength of cores drilled in the direction of casting (vertical cores) will be different than cores drilled at 9 degrees to the direction of casting (horizontal cores). A commonly used approximation based on experimental data is that the strength of a core taken in the horizontal direction will be 8 percent less than the strength of a core that was drilled vertically [ 19]. Although no correction factor is given in the ASTM C42 Standards, this figure has been adopted by the British Standards [19]. All cores in this investigation were drilled in a horizontal direction, therefore this factor was ignored when comparing core test results. Concrete strength varies vertically due to water gain effects with the weaker concrete near the top of the structural member. It is difficult to obtain a general model for this effect because factors such as mix composition, aggregate type, workability, environment and compaction are involved. Results of several studies showed that strength varies linearly from top to bottom and the strength difference between the top and bottom surfaces increases with increasing depth of concrete [19]. Based upon prior 44

58 experimental data, a strength variation of 2 percent was assumed for the bridge girder of this investigation which had a depth of 45 in. [19]. Because the cores were drilled at various heights in the web, it was necessary to correct all the strengths to a reference height on the girder in order to compare them with each other. This was done by using a linear variation in strength of 2 percent from the top of the girder to the bottom of the girder and calculating a factor to correct all cores to mid-depth of the girder (8 in. from the bottom of the web). The twenty approximately 3.8 in. by 7.5 in. cores taken from the bridge girder were tested in an air dry condition in a loading frame at a constant stroke rate. Constant stroke is not necessarily the same as constant strain because the loading machine platens will also deform during loading. Cores were loaded at a rate of 15 sjmm for the initia.l portion of the loading and then at a rate of 25 sjmm to failure. Peak loads were recorded and load versus platen movement curves were obtained. Table 12 gives measured core strength, and the factors for correcting measured strength to strength at mid-depth of the girder. The depth corrected strengths are again shown in Table 13 as well as core length, diameter, L/d ratios, ASTM C42 L/d correction factors and L/d (and depth) corrected strengths. Since L/d ratios approached 2, L/d correction factors were 45

59 either not required or very close to 1. Results from core 11-1 were not available due to an error in operation of the testing machine. The results from the core tests varied greatly from a minimum of 365 psi to a maximum of 9656 psi. The low values obtained were determined to be caused by poor capping practices. ASTM Standard C42 specifies that neither end of compressive test dicularity by more 1/8 in. in 12 in.) specimens should depart from perpenthan.5 degrees (approximately equal to [ 2 ] Based upon the maximum difference between length measurements of the cores taken at four locations around the perimeter of the cores, cores 1-1, 2-2, 5-2, 8, 9-2, 1-2, 11-2, and 12-2 exceeded this limit. These cores are shown in bold print in Table 13. Seven of these 8 cores account for the lowest compressive strengths observed. In addition, upon examination of the cores after failure, most of these cores had broken caps. This was indicative of the caps not having been parallel. Excluding the 8 cores given above, the range of compressive strengths still varied a great deal from 577 psi to 9656 psi. It was possible that some of these 11 remaining cores did not conform to the ASTM C42 specification but this was not detected with the measuring technique used. For example, the core faces may have been parallel to each other but not perpendicular to a line through the center of the core. This may have been the 46

60 case if the drill bit "wandered" when drilling the core. Therefore, the remaining 11 cores, especially those with lower compressive strengths may also be suspect. Failure modes of cores 4-2, 5-l, 6-1, 7, and 9-1 (strengths equal to 577, 671, 6788, 777 and 739 psi, respectively) indicated improper failure with one side of the core failing prematurely. After these results were discarded, then the lowest compressive strength of the six remaining cores was 8222 psi. The average compressive strength for these six cores was 8683 psi with a coefficient of variation of 6.7 percent. In addition, of these 11 cores, 1-2, 5-l, and 6-1 (strengths equal to 8222, 671, and 6788 psi, respectively) contained steel which may have affected the results Two Inch Cores B.ecause tests of the 4 in. cores gave results which were suspect, some of the cores that had been used previously for break-off testing were compression-tested to obtain additional data. The break-off cores were approximately 2.13 in. in diameter and had one formed edge from the surface of the beam web and one rough edge along the break-off fracture surface. Only those cores which had a fairly good perpendicular fracture surface (an L/d ratio of approximately one or greater after surface preparation) and contained no 47

61 embedded steel were selected for testing. In order to avoid the problems encountered previously with the 4 in. nominal diameter cores, the surfaces of the smaller cores were carefully prepared by cutting and grinding to achieve perpendicularity and planeness to.1 in. The 2 in. nominal diameter cores were tested approximately five days after the surfaces had been prepared to allow sufficient time for the concrete to dry under ambient conditions. As for the 4 in. nominal diameter cores, testing was conducted in a load frame using stroke control. Six cores were loaded at a rate of 15 s/mm. The remaining nine cores were loaded at a slightly slower rate of 25 s/mm. Peak load and load versus platen movement curves were obtained. Again, the effect of a smaller core diameter relative to aggregate size as well as the effect of a nonstandard L/d ratio had to be considered. Core length after surface preparation varied from in. to in. and diameters were in the range of 2.13 in.; therefore, L/d ratios were in the range of 1. Munday and Dhir tested six cores each of 2, 3, and 4 in. diameters having a maximum aggregate size of. 79 in. with L/d ratios of 2 and found no difference in measured strength. This led them to conclude that for smaller cores no correction for diameter is required for cores in the range of 2-6 in. diameter having maximum aggregate size of in. [19]. This conclusion was 48

62 supported by a larger study with 2-4 in. cores,.39 in. maxim urn a g greg ate s i z e and an L / d ratio of 1 [ 1 9 ]. AS T M Standard C42 specifies that core diameter should be at least three times the maximum aggregate size [2]. The maximum aggregate size of the bridge girders was 1 in. and the core diameter was approximately 2.13 in. resulting in a core diame ter to aggregate size ratio of 2.13 compared with a minimum ratio of 2.53 as allowed by Munday and Dhir above. For L/d ratios around 1, correction factors given in the literature range from approximately.75 to.9. Although it has been shown that correction factors increase (thus less of a correction) as strength of the cores increase, both the British Standards and ASTM Standard C42 give a single average correction factor for an L/d ratio of 1 (.8 and.87 respectively) [2, 21]. Kesler [22] and Sangha and Dhir [ 19] determined correction factors as a funct~on of measured strength. These relationships are shown in Figure 31. Swamy and Al-Hamed [23] performed compression tests upon 2 in. diameter cores with concrete strength, L/d ratios (1 or 2), and coarse aggregate size (.39 and.79 in.) as variables. Their resulting correction factors for.39 in. aggregate are plotted in Figure 31. They also found that concrete with larger aggregate (.79 in.) required less correction (a greater correction factor) than the s arne concrete mix with.39 in. aggregate. This correction factor 49

63 was found to be approximately 13 percent greater for crushed gravel aggregate of a 41 psi strength. They did not determine how this difference in correction factor varied at higher strengths. The coefficient of variation was greater for larger aggregate concrete (11.6%) than for the smaller aggregate concrete (7.6%). However, the coefficient of variation showed a consistent decrease with increasing strength for all concrete mixes tested although this was only proven for the smaller aggregate concrete because only the smaller aggregate concrete was tested at different strength levels. Based upon the strength-correction factor relationships discussed above, and in view of the lack of data available concerning the behavior of small high strength cores tested in compression, it was determined that ASTM C42 correction factors would be sufficiently accurate in the strength range requir.ed to use them for the 2 in. break-off cores. These factors are more conservative (result in a lower strength) for the strength range in question (around 9 13 psi measured strength) than either the Kesler, Sangha-Dhir, or Swamy-Al Hamed factors. Measured core strength, distance in inches from the bottom of the web, and the height-corrected measured strength for the 2 in. cores are given in Table 14. As for 5

64 the 4 in. cores, core strength was corrected to mid-depth of the beam by assuming a 2 percent variation in concrete strength between the bottom and top of the girder. Measured (depth-corrected) core strengths were corrected for L/d ratios using ASTM C42 factors. These values are given in Table 15. For an L/d value of exactly 1. the correction facto-r used was. 87. At other values of L/d, correction factors were obtained by linear interpolation between the recommended ASTM C42 factors. Corrected compressive strength averaged 1216 psi with a coefficient of variation of 8.54 percent. Without correcting for height, L/d corrected compressive strength averaged 118 psi with a coefficient of variation of 8.7 percent. Figure 32 compares 2 in. core strengths that have not been corrected for height (denoted by a cross on the graph) with height-corrected strengths (denoted by a box). In this figura, boxes and crosses located at the same x-coordinate (the distance from the edge of the beam) represent corrected and uncorrected values for the same core. Correction factors for L/d ratio have been applied to both sets of data in this figure. Figure 33 shows the corrected measured strengths (corrected for both height and L/d ratio) for the 2 in. cores in relation, to their location along the length of the beam. These values are the same as denoted in the previous 51

65 figure by the boxes; but in that figure, the data points were plotted on an expanded y-axis (strength) scale. The mean value of the strength for the fifteen 2 in. cores after correcting for height and L/d ratio as per ASTM C42 was 1216 psi, the standard deviation was 872 psi and the coefficient of variation was percent. The mean concrete strength of the population is given by: (Eqn. 18) where ta,n- 1 is the value of the t distribution (used when the sample size is less than 3 see Appendix A for description of the t distribution) at confidence limit a for a sample with n-1 degrees of freedom and m is the sample mean. The resulting mean is: Taking a as 95 percent, ta,n- 1 is and JJ. is within the range, 9733' psi ~ JJ. ~ 1699 psi. The coefficient of variation was slightly lower for the height corrected strength than for the uncorrected strength (8.54 versus 8.73 percent). There was no significant difference between the strengths of the break-off cores in 52

66 the cracked and uncracked sections of the beam although the coefficient of variation was higher for the cracked beani section (9.25 versus 7.25 percent) Core Compression Tests 2nd Beam Because of conflicting compressive strengths given by the two series of core compression tests (approximately 12 psi for 2 in. cores and 868 psi for 4 in. cores), additional cores were drilled and tested to obtain additional data. Six 2 in. nominal diameter and six 4 in. diameter cores were drilled from the web of a second identical girder of the bridge. All cores were obtained from uncracked sections at either end of the second beam after it had undergone load-testing. Cores were obtained using the same drill and drill bit as before and drill times were compar.ab le. Care was taken when setting up to dri 11 each core that the drill bit was perpendicular to the beam to obtain cores with surfaces at right angles to the sides. The cores were tested in an air dry condition in a load frame using load control at loading rates comparable with previous load rates. Half of the six 4 in. diameter cores were capped with sulfur cement such the surfaces were plane to at least.2 53

67 in. (However, one core, core 4-1 was plane to.1 in. The other two cores were plane to.15 in. and.2 in.) The ends of the remainder of the cores were prepared by grinding them plane to within.1 in. L/d ratios were approximately 2. for the three capped cores. The remaining three cores had L/d ratios of 1.851, 1.77, and Table 16 gives measu red strengths for the 4 in. cores, corrected strength to mid-depth of the girder, ASTM C42 L/d correction factors, and corrected strength. Average strength (corrected for height and for L/d ratio per ASTM C42) for this set of cores was 842 psi. This was approximately 64 psi less than the ave rage of the six cores determined to be valid from the first be am ( psi). The coefficient of variation from the second series of tests (15.6 percent) was high compared with the previously measured coefficient of variation (6.7 percent) and the expected within test variation (6 percent for 3 cylinders [ 24]). The high coefficient of variation resul t.ed from the three cores with L/d ratios of approximately 2 having uncorrected compressive strengths that ranged from the highest to the lowest compressive strengths measured for this set of cores (643 to 9688 psi). For this group of cores, testing of core 4-6 with the lowest L/d ratio (1.172) resulted in the highest measured (uncorrected) strength (185 psi) as expected. However, the cores with the next two highest measured strengths (core 4-1: strength = 9688 psi and core 4-2: strength = 8335 psi) actually had 54

68 L/d ratios of approximately 2, higher than the remaining cores. These two cores had capped ends rather than ground ends. Core 4-1 was plane to a very good tolerance of.1 in. (The other two cores had tolerances of.15 and.2 in.) The capped cylinders resulted in both the highest and the lowest corrected concrete strengths for these six cylinders. The number of samples was insufficient to determine if any difference existed for capped-end versus ground-end cores, but it appears that capping high strength cores contributes to the variability of the results. This is illustrated in Figure 34 which shows the measured strengths of both the 4 in. and 2 in. cores from the second beam plotted against L/d ratios. On this graph points labelled 4-1 through 4-6 represent 4 in. cores and points labelled 2-1 through 2-6 represent 2 in. cores. The capped-end cores (those with L/d's of around 2) are represented by the triangles. Table 17 gives results for the two inch cores from the second beam. All of the 2 in. cores had ends prepared by grinding. L/d ratios for the 2 in. cores ranged from.994 to The average (height and L/d ratio) corrected compressive strength was approximately 97 psi, 5 psi less than the previously obtained average for 2 in. cores of 12 psi. The coefficient of variation was 8. 7 percent, comparable with the coefficient of variation for the other series of 2 in. cores (8.5 percent). 55

69 Compression tests of cores from both beams resulted in 15-2 percent greater strengths on average for the 2 in. cores than the 4 in. cores (based on using AS TM C42 L/ d correction factors). The average compressive strength (in psi) and coefficients of variation (COV) for each of the two beams and for the two sets of data combined are summarized below: 1st beam 2nd beam Combined 4 in. cores 8683 (6.7% COV) 842 (14.6% COV) 8363 (11. 8% COV) 2 in. cores 1216 (8.5% COV) 9693 (8.7% COV) 167 (8.7% COV) This difference in strength was unexpected as research has shown that core diameter only affects coefficient of variation but not compressive strength and L/d effect should have been corrected for by the ASTM C42 factors. Possible explanations are l)the aggregate size to core diameter ratio was greater than allowable, thus affecting results, or 2)ASTM C42 L/d corrections used were not applicable to small diameter cores. One explanation for the difference between strengths of the 2 and 4 in. cores is that the different end preparation methods affected measured strength, however this was not verified by the 4 in. cores for the second set of data. For 56

70 the first beam, the two inch cores had ground ends and the four inch cores were capped. Werner [25] found compressive strengths for strong (approximately 7 psi) cylinders capped with sulfur cement to be from 8. 3 to 19.7 percent less than strengths obtained with aluminous cement caps (which gave the greatest compressive strength). There was littre difference between results using different capping materials for weak (2 psi) concrete. This effect can be attributed to the fact that the lateral expansion of the capping material may be different than the lateral expansion of concrete. For a weak, soft capping material with a high ratio of Poisson's ratio to modulus of elasticity, the expansion of the capping material may be greater than the expansion of the concrete thus increasing the lateral strain in the concrete and resulting in a lower apparent concrete strength. Conversely, a strong capping material will res train lateral expansion and a higher apparent cone rete strength will be obtained. Testing machine platens may also restrain the specimen, depending upon the amount of friction between the test specimen and the platen [26]. The average test results for the two beams were fairly close for both the 2 in. cores (5.2 percent difference) and the 4 in. cores (7.7 percent difference) although the cores were drilled and tested by different individuals. 57

71 Regression analysis was performed using 2 and 4 in. core data from both beams in order to calculate L/d correction factors. Regression analysis using a power model resulted in the following relationship: f' c (L/d) -o 5191 (Eqn. 19) The correlation coefficient, r, was ) indicating a good correlation. The regression curve and the data points are shown in Figure 35. The value of f'c at L/d 2 was 8128 psi. Based on the regression equation, L/d correction factors were calculated by dividing the value of the regression equation at L/d of 2. by the value of the equation at the L/d ratio in question: factor [ 2/(L/d)] -o (Eqn. 2) The calculated L/d correction factors based on the regression equation compared to ASTM C42 and BSI factors are shown in Figure 36. The correction factors calculated were significantly smaller (a greater correction) than either the ASTM C42 factors or the BSI factors. This is contrary to the fact that a higher strength concrete will require less correction than a lower strength concrete (see Section 3.2.6). Therefore, the difference between the correction factors is most likely due to either the small core dia- 58

72 meters or the large aggregate size to core diameter ratio for the 2 in. cores. When core strengths were corrected using Equation 2 instead of the ASTM correction factors the following results were obtained: 1st beam 2nd beam Combined 4 in. cores 8614 (5.8% COV) 7676 (14.% COV) 8145 (12. 3% COV) 2 in. cores 8147 (7.4% COV) 8237 (6.% COV) 8173 (7.2% COV) The data based on the calculated L/d correction factor indicates an average compressive strength for the concrete of approximately 815 psi. This value, as well as the higher value of approximately 1,2 psi obtained from 2 in. cores using ASTM C42 correction factors, is compared with results of non~estructive tests in the following chapter. 59

73 4. Data Analysis In the following sections, results of the core compression tests are discussed and the nondestructive test data are compared with results of the core compression tests. Because of the small number of sample points for 4 in. diameter cores from the first beam due to problems with capping methods and materials, compression tests of 2 in. diameter break-off cores were used in this analysis. The compression test results (corrected for L/d by both ASTM C42 factors and Equation 2) were compared with the correlation relationships from the literature presented in Chapter 2. Appendix A outlines the regression analysis methods utill.zed Concrete Compressive Strength from Core Tests The average 28-day compressive strength of the concrete cylinders for the girder of this study was 67 psi. The average strength after 2 years based on compression tests of the 2 in. cores that were corrected for L/d ratio per ASTM C42 was approximately 12 psi or a 52 percent increase in compressive strength between 28 days and 2 years. If correction factors calculated by Equation 2 are used, then the average compressive strength for the girders 6

74 was approximately 815 psi or a 22 percent increase in compressive strength over the time period. Washa and Wendt [27] observed concrete compressive strength gains ranging from 27 percent to 37 percent for the period between 1 month and 25 years for concrete made in 1937 from high early strength cement. For standard cement concrete, the strength incre~se for the same time period ranged from 57-9 percent. The cylinders were stored outdoors and subjected to about 25 freeze-thaw cycles each year. In general, they found strength gains occurred from 1 month to 1 years with moderate decreases in strength between 1 and 25 years. The 52 percent increase observed for ASTM C42-corrected cores in the case under study was between Washa and Wendt's results for high early strength cement concrete and their results for standard concrete. The results based on the derived correction factor correspond more closely to Washa and Wendt's results for high early strength concrete. Of course, these.comparisons are based on the assumption that the 28- day standard concrete cylinder tests for the bridge girder accurately represented the girder concrete strength in situ at that point in time. 61

75 4.2. Rebound Number - Strength Relationships for 2 in Cores Corrected per ASTM C42 For each 4 in. core location, 18 rebound readings were obtained in situ as described in Chapter 3. Only comparisons based on 2 in. cores corrected for slenderness effects as per ASTM C42 are discussed in this section. Results based upon the calculated correction factor from Equation 2 will be discussed in the following section. The average rebound readings were correlated to the compressive strength of the 2 in. break-off cores (ASTM C42-corrected) from the same test area. However, the rebound hammer tests and break-off cores were not at the same vertical location on the web and could not be compared directly due to the variation in concrete strength from top to bottom of the girder caused by water gain effects. In order to be able to compare the break-off core strengths with the rebound hammer readings, break-.off core strength readings were corrected to the height of the 4 in. cores (where the rebound readings were taken) assuming a 2 percent linear variation in concrete strength from top to bottom of the girder. Test areas had one or two 4 in. cores for which rebound readings were correlated with strength values from one, two, or three breakoff cores. The average of the break-off core compressive strength results that had been previously corrected to midheight of the beam were adjusted to the height of the 4 in. 62

76 core (or the average height of the two 4 in. cores in an area) In this manner the corrected average break-off core compressive strength could be compared with the average rebound hammer reading for each test area. Linear regression analysis was performed on this data and the results are shown in Table 18. The regression equation obtained from the analy-sis was: f' c 246.7R (Eqn. 21) where R is the rebound number. The value of r 2 was.26; therefore, 26 percent of the variation in f' c could be attributed to variations in R. An r 2 of. 26 for 8 data points indicates a weak linear correlation. The hypothesis that a linear relationship does not exist was tested following the procedure outlined in Appendix A. The hypothesis was rejected at a.2 (2 percent) level of significance or gr~ater and was accepted at levels of significance less than.2. The level of significance is defined as the probability of rejecting the hypothesis when it is true. The 9 percent confidence interval on a future observed response at an R value of 54 ranged from psi to psi or ± 14.4 percent of the estimated mean. Figure 37 shows the observed data points (ASTM C42-corrected), the regression line, and the 9 percent confidence intervals. Regression 63

77 analysis utilizing a power curve model gives no better results with r 2 also equal to.26. Using Facaoaru' s method of calculating a coefficient, Ct, for concrete other than the non- reference concrete gave a Ct value of 1.95 and resulted in the following equation: f' c 1.59 R 2 2 (Eqn. 22) where R is the rebound number. Ct was calculated by dividing the average 2 in. core compressive strength at mid-height (1191 psi, ASTM C42-corrected) by the calculated value of f'c from Facaoaru's equation (Equation 2) using the average rebound number for the corresponding cores (53. 78) as the value for R. Figure 38 compares the linear regression equation obtained from the regression analysis with correlation relationships developed by the rebound ham~er manufacturers and by Facaoaru (for the reference concrete) [ 6]. At the higher values for rebound numbers, the manufacturer's curve gives significantly lower strengths than observed in this study (for ASTM C42 L/d corrected strengths). Facaoaru's equation estimated the concrete strength better than the manufacturer's equation. Table 19 shows the difference between the 2 in. core compressive strength and compressive strength calculated by Facaoaru's and the manufacturer's equations. 64

78 4.3. Rebound Number Strength Relationships for Cores Corrected per Equation 2 Regression analysis was performed with the same data as in the previous section but using correction factors calculated by Equation 2 rather than ASTM correction facto~s. This resulted in the following equation: f' c 186.3R (Eqn. 23) with a value of r 2 of.337. This value, although indicative of a slightly better correlation than in the previous section, still corresponds to a weak linear relationship between strength and rebound number. A linear relationship could not be said to exist at significance levels of less than.15 (15 percent) (see Appendix A). Linear regression analysis using 4 in. core strength (6 data points) versus rebound number for the cores resulted in no correlation (r 2 =.6). Based on the data using the calculated L/d correction factors, Facaoaru's factor, Ct, for the girder concrete was.87. This factor was obtained by dividing the average 2 in. core strength at mid-depth of the girder (corrected for L/d effects per Equation 2) by the value of f'c ref from Equation 2 based on the average rebound number. The 2 in. 65

79 core strength corrected as per Equation 2 compared well with the rebound hammer manufacturer's strength/rebound number relationship (Equation 1). For the average rebound number of 53.8, the manufacturer's relationship gives a concrete compressive strength of 7919 psi. The average corrected compressive strength was 8128 psi, a difference of appro ximately 2.6 percent. Table 2 compares 2 in. core compressive strength with compressive strength by the manufacturer and Facaoaru' s equations (Equations 1 and 2). The value shown for Facaoaru' s equation is based on the reference concrete without using the Ct correction factor. Figure 38 gives the data points corrected per Equation 2 as well as the points corrected per ASTM C42 and the correlation relationships Pulse Velocity - Strength Relationships Because of the relative lack of sensitivity of the pulse velocity measurements in relation to the range of concrete strengths observed, no relationship between pulse velocity and break-off core compressive strength was obtained from a regression analysis. Table 21 gives the average pulse velocity after coring and the corresponding 2 in. core strengths for each test area. Pulse velocity readings were taken through the 4 in. cores both before and 66

80 after the cores were drilled. The pulse velocities shown in Table 21 represent averages for each test area after the cores were drilled. Two in. core strengths corrected for L/d effect by both ASTM C42 and Equation 2 are also given in Table 21. The core strength results had to be adjusted to correct for the fact that they were not at the same height as the 4 in. cores. Core strengths for each area were adjusted to the average height of the 4 in. cores in each test area using the same method described in Section 4.2. As discussed in Section 3.2.4, the pulse velocity measurement sensitivity was no better than ±2 ft/s. For all 2 pulse velocity readings, the sample mean was 15,1 ft/s and the standard deviation was 24 ft/s. Using Equation 18 to calculate the population mean at 95 percent confidence limits results in the population mean pulse velocity 15, ft/s ~ p ~ 15,2 ft/s. None of the pulse velocity values in Table 21 varied by more than 2 ft/s from the probable population mean. Thus the pulse velocity values were not significantly different from each other and a regression analysis would not be valid. From Section 3.2.4, strength values are only accurate to +589 psi or -559 psi based upon the average measured pulse velocity and the theoretical pulse velocity correlation (Equation 9). 67

81 Figure 39 shows the ASTM C42-corrected data points for the pulse velocity/strength relationship and the population mean ranges for pulse velocity and compressive strength (9749 psi ~ ~ ~ 1683 psi for strength and 15 ft/s ~ ~ ~ 152 ft/s for pulse velocity). Also shown are the sensitivity ranges based on the accuracy of the measurements. All the data points were either within or very nearly within the boundaries shown in Figure 39, illustrating how the pulse velocity measurements were not sensitive enough to detect differences in concrete strengths over the range tested in this study. Pulse velocity sensitivity can be improved by measuring pulse velocities over longer path lengths or by using transducers of higher vibrational frequencies. The average pulse velocity reading for all cores included in the analysis was 1511 ftjs. This corresponds to an average strength of 1185 psi based on the theoretical relatlonship with K=.85 (Equation 9). This value is 9 percent greater than the average height-corrected, ASTM C42- corrected 2 in. core strength of psi and 36 percent greater than the average strength corrected by Equation 2 of 8128 psi. Figure 4 (and Table 21) shows the strength of the 2 in. break-off cores in relation to the pulse velocity readings for each test area compared with the pulse velocity/strength relationships discussed in Chapter 2. The theoretically derived pulse velocity/strength relationship 68

82 (Equation 9) with a K-value of.85 better predicted compressive strength on the average than the same equation using a higher K-value based on ASTM C42-corrected strength. A K-value of.85 corresponds to a Poisson's ratio of.24 (from Equation 6). Solving for K in Equation 9 using the average pulse velocity reading (1511 ft/s) and the average ASTM c42-corrected 2 in. core strength (123.5) results in a K-value of. 82 which corresponds to Poisson's ratio of. 26. This value is nearer the suggested dynamic Poisson's ratio of.24 [8], than the normal (static) Poisson's ratio for concrete of.2 [11]. However, solving for K based upon the strength corrected by Equation 2 (8128 psi) resulted in a K-value of. 73 and an even greater Poisson's ratio of.31. Using Facaoaru's relationship for pulse velocity and strength requires the calculation of a coefficient, Ct, to correc.t for a concrete other than the reference concrete [6]. This coefficient was calculated by dividing the average 2 in. core strength by the strength calculated using the average pulse velocity value in Equation 1. For the ASTM C42-corrected cores, this resulted in a Ct value of 1.89 and f'c was then given by: f' c (l. 8 9) ( 34. 8)e(3.35xlOE-4V) (Eqn. 24) 69

83 This equation is also plotted in Figure 4. Although Ct for rebound testing and Ct for pulse velocity testing will not necessarily be the same for the same concrete, the Ct value of 1.89 (ASTM C42-corrected strengths) obtained above deviated a great deal from 1. while the Ct value obtained from rebound testing for ASTM C42-corrected strengths was nearer 1. (1.95). For the ASTM C42-corrected cores with Ct taken as 1., Facaoaru's equation for rebound testing was much better at predicting compressive strength than Facaoaru's equation for pulse velocity. For pulse velocity testing, the theoretical equation using a K-value based upon the dynamic Poisson's ratio of concrete was the best predictor of concrete compressive strength. For compressive strengths corrected by Equation 2, the value of Ct calculated by Facaoaru's pulse velocity/strength equation was The Ct value calculated for these cores by FaGaoaru's method for rebound hammer was.87. For both the ASTM C42-corrected strengths and the strengths as corrected by Equation 2, the theoretical equation with a K value of.85 overestimated strength and Facaoaru's equation for the reference concrete (without the Ct correction factor) underestimated strength (see Table 21). Combined rebound hammer-pulse velocity/strength models with rebound hammer and pulse velocity as the independent 7

84 variables and concrete strength as the dependent variable (Equations 12, 13 and 14) could not be developed from the data obtained in this investigation. This inability to develop a good combined model was due to the difficulties encountered in developing a relationship between pulse velocity and strength alone. Models using Equations 12 and 14 resulted in a negative coefficient for the pulse velocity term (an inverse pulse velocity strength relationship). The model based on Equation 13 resulted in a for the pulse velocity term indicating a small coefficient negligible pulse velocity contribution to the equation Windsor Probe - Strength Relationships Table 22 shows average Windsor probe readings and corresponding 2 in. core strengths for each area. Core strength corrected for L/d effects by both ASTM C42 and by Equation 2 are shown in the table. Core strengths were corrected to the average height of the Windsor probe readings based on the same assumptions discussed in previous sections of this text. Figure 41 shows a plot of this data compared with the manufacturer's correlation equations. No trend indicating a relationship between probe length and compressive strength was apparent within the range of strengths observed in this study. However, it is clear that 71

85 the manufacturer's equations underestimated compressive strength for strengths corrected by both methods. The average exposed probe length for the areas for which core compression tests were performed was 2.18 in. and the corresponding average corrected 2 in. core strengths were 123 psi (by ASTM C42) and 811 psi (by Equation 2). The manuf-acturer's equation (Equation 16) predicted a compressive strength of 684 psi (aggregate Moh's hardness of 6). This was lower than the average measured compressive strength (see Table 22) Break-off Test - Strength Relationship Table 23 shows break-off test results and corresponding core compression test results (results were corrected for L/d effects by the two different methods discussed in Chap te_r 3) for each 2 in. core tested. Break-off test results exhibited a high standard deviation and did not compare well with manufacturer's strength correlations for the break-off equipment (see Section ). No relationship was observed between core strengths (neither the 2 in. cores corrected per ASTM C42 nor the cores corrected per Equation 2) and break-off test results. This was attributed to the large aggregate size (1 in.) and possible damage to 72

86 the cores during the drilling operation from lateral vibrations of the drill bit Concrete Uniformity -The coefficient of variation (COV) for each of the tests is given below and compared with within test coefficients of variation from the literature. The number of tests corresponding to each COV is given in parentheses: cov Within Test cov 4 in. core test (ASTM C42) 6.7% ( 6) 6% ( 3) [ 24] 4 in. core (Eqn. 2) 5.8% ( 6) " 2 in. core test (ASTM C42) 8.5% (15) 11.6% (8) [ 2 3] 2 in. core (Eqn. 2) 7.4% (15) " Rebound hammer 3.% (21) 7.4% (18) [ 1] Pulse.velocity 1.3% (2) 2.% [ 7] Windsor probe 3.6% (12) % [ 2] Results from break-off test are not included because of invalid test results (see Section 4.5). The core compression test results given are from the first beam corrected for L/d effects by both methods discussed in Chapter 3. The coefficient of variation for 4 in. core tests given is for the six valid tests (see Section 3.2.6). The COVs given for 73

87 rebound hammer, pulse velocity, and Windsor probe are for tests performed at all core locations. The value given as the within test COV for pulse velocity (2%) is not actually a coefficient of variation but the maximum variation observed in a series of tests. This data combined with the inability to obtain good regression equations for the nondestructive tests indicates that the concrete strength was relatively uniform along the girder in relation to the accuracy of the test methods used. In Figure 22, the highest rebound reading and the lowest rebound readings at a particular location follow similar trends along the length of the beam. The rebound hammer, Windsor probe and pulse velocity test results along the length of the beam are compared in Figure 42. Plotted against location along the length of the beam are: from top to bottom, average exposed Windsor probe length, rebound number, (before and after beam load testing), and pulse velocity (before and after drilling cores) For the three tests, similar trends are apparent for the area from approximately 16 to 24 feet with decreasing concrete strength toward the center of the span. Otherwise, no trends are readily apparent when comparing the tests. 74

88 5. Corrosion of Steel in Concrete The use of deicing salts to maintain ice-free roadway surfaces in areas of the United States where ice and snow are a problem during the winter months became an increasingly common practice beginning in the mid-195's. In the United States, the amount of salt used on the highways increased from one million to ten million tons per year from 1955 to 1976 [28] Severe deterioration of many concrete bridge decks resulting from the use of deicing salts was observed by the late 196's in the form of spalling and disintegration of the concrete cover. Because the deicing salts are directly applied to bridge decks and roadway surfaces, most of the damage observed to date has primarily affected these surfaces. The bridge superstructure is exposed to the salts primarily from mist and spray created by vehicles travelling beneath the bridge. Salt water from the bridge deck may also drip through joints. The magnitude and effect of salt exposure on the bridge superstructure is of primary interest because of its possible effect upon s true tural integrity. This study provided an opportunity to measure the amount of salt (chloride ion) penetration into the bridge girders of a highway overpass that had been in service over a twenty year period from the mid-196's to the mid-198's. The amount of salt penetration in the bridge 75

89 deck was also measured and compared with chloride ion levels measured in the bridge girders Background -In general, the major problem caused by corrosion of ordinary reinforced concrete is not reduced strength of the steel but cracking and spalling of the concrete caused by the corrosion products. These corrosion products occupy more volume than the steel from which they formed; thus these corrosion products exert mechanical stresses on the surrounding concrete and may cause cracking and spalling. In prestressed concrete, however, because of the high stress levels in the steel, the greater concern is the loss of steel cross section and thus the reduction of load bearing capacity of the prestressing steel. In addition to generalized corrosion and pitting resulting in reduced surface area of the prestressing steel, other corrosion mechanisms of possible concern in prestressed concrete include stress corrosion cracking and hydrogen-induced cracking. Also of concern is the effect of a corrosive environment on the fatigue strength of prestressed concrete. These different corrosion mechanisms for prestressed concrete will be discussed after outlining the environmental factors that 76

90 play a role in the generalized corrosion of steel in concrete. The total time to corrosion of reinforcing steel in concrete is the sum of the time necessary for the conditions at the reinforcing steel to be conducive to corrosion plus the time for the corrosion to proceed to a point where repair of the structure is required. In the case of chloride induced corrosion of steel in concrete, a threshold level of chloride ions in the concrete environment at the level of the reinforcing steel is necessary for the initiation of corrosion. There have been conflicting reports of the threshold chloride ion concentration that will cause corrosion of steel in concrete. These will be discussed in subsequent paragraphs. The properties of concrete that affect the onset and the r~te of corrosion include its relatively high ph and its permeability. Recent studies have shown that uncontaminated Portland cement concrete has a ph between 13.5 and At this high ph, steel will not corrode unless something such as the chloride ion or carbon dioxide (discussed later) intrudes to change the protective nature of the concrete environment [29]. 77

91 The corrosion rate of steel in aerated water solutions of ph 4 through 12 is 1 mils/year. At the ph range 1 through 13, the corrosion rate decreases by an order of magnitude over the range such that the corrosion rate at ph 13 is approximately 1 mil/year [28] In high ph solutions, a protective film of gamma-fe 2 3 (ferric oxide) is formed on the steel which acts as a barrier against corrosion. However, the presence of the chloride ion reduces or destroys the protective nature of this passive layer and leads to accelerated corrosion or pitting attack. The method by which the chloride ion destroys the passivity of the protective layer has not been fully established [29]. Permeability of the concrete as well as the presence of cracks is of importance because the chloride ion must diffuse through concrete to reach the steel for the corrosion process to be initiated. Oxygen is required to support the c.orro s ion of s tee 1 in concrete and must also diffuse through the concrete to reach the level of steel. Permeability of concrete is related to its water/cement ratio. A higher water/cement ratio will lead to larger pores or a greater number of pores thus increasing the permeability of the concrete. In ASTM Special Technical Publication 818, Slater reviewed the research concerning the threshold chloride ion 78

92 concentration necessary for initiation of corrosion of steel in concrete [28]. The conclusions of those studies are discussed below. Berman [3] and Hausmann [31] in separate studies used saturated aerated Ca ( OH) 2 (calcium hydroxide) solutions (ph 12.6) to simulate the high ph environment of concrete and found the chloride threshold for steel corrosion in the Ca(OH) 2 solution was 7 to 1 ppm chlorides. Berman rep laced oxygen in the Ca ( OH) 2 solution with nitrogen and the threshold concentration of chloride ion for the initiation of corrosion increased greatly, proving the critical role that oxygen plays in supporting the corrosion reaction. Hausmann immersed steel in NaOH (sodium hydroxide) solutions of varying ph and found that the threshold chloride level for corrosion was dependent upon ph of the solution. As the ph of the solution was increased the threshold level of chloride ions necessary for initiation of corrosion also increased [31]. Instead of studying steel immersed in a high ph solution, a more direct method of studying corrosion of steel in concrete involves soaking small slabs or prisms with embedded reinforcing bars in a salt water solution. Using this method, average threshold values of 25 to 51 ppm chlorides by weight of concrete have been observed by investigators [32, 33, 34]. In order to convert chloride ion concentration by weight of concrete to chloride ion concen- 79

93 tration by weight of cement, the cement factor of the concrete must be known. For the bridge girders, the cement factor was 2 percent by weight of concrete. Based on this, the range of threshold chloride ion concentrations given above is equivalent to 125 to 255 ppm chloride ion by weight of cement. The threshold level of chloride leading to corro~ion of steel in an aqueous Ca(OH)2 (calcium hydroxide) solution of 7 to 1 ppm is lower than that observed in concrete (125 to 255 ppm by weight of cement). One theory for this difference is that a lime-rich layer formed on the surface of the reinforcing steel in a concrete environment provides protection against corrosion in addition to the protection provided by the high ph of the concrete environment previously discussed [29]. Studies of bridge decks in service have provided additional data on threshold chloride concentrations. A survey by Van deveer of 473 bridge decks showed that chloride levels of 4 ppm by weight of cement were associated with corrosion [ 35]. The Federal Highway Administration has set 3 ppm by weight of cement as the chloride level at which concrete of a bridge deck must be replaced. The deck can be left intact if the chloride level is 15 ppm by weight of cement or less and further investigation is required to determine action for chloride levels between 15 and 3 ppm [ 2 9]. 8

94 The factors affecting chloride-induced corrosion of steel in concrete as relating to the reduction of load bearing area of the prestressing steel have been discussed in the previous paragraphs. Another method by which corrosion may be initiated involves the penetration of carbon dioxide into concrete. Carbon dioxide will react with Portland cement to produce carbonates. The ph of the concrete is lowered and corrosion may result. Carbon dioxide penetration rates are dependent on the permeability of the concrete. However, carbonation rates for good quality, dense concrete are low and chloride ion penetration is the primary factor in determining time to corrosion of concrete structures exposed to deicing salts in the United States. The gene rally accepted relationship be tween carbonation depth and time is given by: Carbonation depth (Eqn. 25) where C is a constant dependent on concrete permeability and environmental conditions and t is time [36, 37] Note that carbonation rate decreases with time. Assuming a moderate strength concrete, carbonation depth would be on the order of 1 to 2 5 mm (. 4 to 1 in. ) after 2 5 years [ 3 7 ]. For high strength concrete the carbonation depth would range between 5 to 1 mm (.2 to.4 in.) after 25 years [37], much less than the required concrete cover for reinforcing steel. 81

95 Also of concern in prestressed concrete is the possibility of unexpected brittle failure of the prestressing steel due to stress corrosion cracking. Stress corrosion cracking is the result of a combined stress and corrosion state. Conventional reinforcing steels have low to medi urn carbon content and are therefore not susceptible to stress corrosion cracking. At low ph values in the presence of hydrogen sulfide (H 2 S), however, stress corrosion cracking can take place in the higher strength steels used in prestressed concrete. In a normal concrete environment at high ph values, stress corrosion cracking will not occur. In a survey of failures of prestressing steel in concrete, the problems encountered appear to be related to general corrosion and not stress corrosion cracking [28]. Griess and Naus tested A416 prestressing strands exposed to different solutions containing chlorides, hydrogen sulfide and ammonium nitrate using the slow constant strain rate techni,que. This method involves tension loading of a test specimen at a very slow strain rate while exposed to the test environment until fracture occurs. By comparing time to failure under different conditions, the relative susceptibility of the test specimens under different environments can be established. They found that no cracking occurred in any chloride solution regardless of ph. These results were in conformance with the general belief that chloride ions do not produce cracking in steels of this type. Cracking was 82

96 observed in solutions containing hydrogen sulfide at a ph less than 7 [38]. Cracking of high-strength steels in the presence of hydrogen sulfide has been attributed to hydrogen embrittlement, one mechanism by which stress corrosion cracking can occur. The hydrogen sulfide facilitates the entry of hydrogen into the steel by interfering with the formation of molecular hydrogen at the cathode area [ 3 8 l. Hydrogen embrittlement can also occur when hydrogen produced in cathodic areas by the corrosion reaction enters the high strength steel. At the high ph values of the uncontaminated concrete environment, however, this type of cracking will not occur [29]. Evidence is available that suggests that exposure of concrete to chlorides reduces its fatigue life. Fatigue strength of steel reinforcement is reduced when bars are tested in saltwater rather than in air. This is true for bare bars as well as for bars with 1 in. cover. Although this data pertains to specimens continuously immersed in saltwater, effects may also be severe for a wet/dry saltwater environment. When reinforced concrete prisms tested in fatigue fail, they fail at cracks in the concrete because of stress concentrations at the crack. This is true when testing in air as well as in saltwater. In saltwater, the 83

97 crack may also provide a path for ingress of the chloride ion, thus accelerating the fatigue process. There are many variables which affect the fatigue behavior of reinforced concrete: Material Type of steel Bar size Protective coatings Concrete quality Concrete cover Construction joints Loading Mean stress Amplitude Frequency Load history Environmental Temperature Cathodic protection Wet/dry The effect of these variables on reinforced concrete has not been well described [28, 39]. Because of the service conditions of the bridge girders in this study, the most likely mechanism of corrosion of the reinforcing steel was chloride-induced corrosion. Although threshold chloride ion concentrations necessary for the initiation of corrosion have not been well established, the lowest concentrations reported to cause corrosion were in the range of ppm by weight of concrete. These values were considered to be the critical chloride ion level for this investigation. 84

98 5. 2. Chloride Ion Concentrations in Bridge Girders The major exposure of bridge girders to deicing salts comes from spray caused by vehicles passing under the bridge. The highest chloride concentrations were expected to be found on the side and bottom surfaces of the facia (external) girders located above the incoming lane of traffic because mist and spray from trucks and automobiles passing under the bridge would be carried toward the bridge by the forward motion of the traffic. As shown in Figure 1, this would be the west face of girder 7 and the east face of girder 16. The faces of interior girders toward incoming traffic would be expected to show higher chloride concentrations than the sides facing departing traffic. The girder upon which the tests described to this point in the study were carried out was an interior girder (either girder: 8, 9, 1, 13, 14, or 15 as shown in Figure 1). Because the interior girders were identical and the girders were not marked to show their original location when they were removed from the b!idge, it was not possible to determine the exact location of this interior girder. Chloride ion concentrations were checked at 2 locations along this beam. Throughout the rest of the text this b earn will be referred to as girder 1 or beam 1. 85

99 Because it was thought that the bridge girder with the highest expo sure to salts would most 1 ike ly be one of the facia girders facing incoming traffic and not an interior girder, points along two facia girders were also drilled and analyzed for chloride ion concentration. One of the facia girders checked for chloride ion levels was located in the structures laboratory at the University of Minnesota to be used in other tests. This girder was referred to as girder 2 or beam 2. It was determined that be.am 2 was the facia girder that was located above the incoming traffic lane because it was apparent from the position of several bolt holes that a traffic sign had been located on the beam face. Samples were drilled at 22 points on beam 2. In addition, a facia girder from the bridge located in the storage yard was also checked for chloride ion levels. It was unknown whether the girder was located over incoming or outgoing traffic. This girder was referred to as girder 3 or beam 3. Seven sampl~ points were drilled from the facia side of girder L Data Collection Procedure At each of the selected test points, sample dust was collected by the following procedure. Using a Milwaukee hammer drill with a 3/4 in. bit, the surface of the concrete was scratched (approximately 1/16 in. depth) and the drill 86

100 dust was discarded. Drilling was then continued at 1/2 in. increments. At each increment the drill dust was collected and put in a sample bag. The drill bit and implements for collecting the samples were cleaned and the hole was vacuumed after each sample was taken. Samples were analyzed by the "Berman method" for determining total chloride content [4]. Chloride ion concentrations were reported in parts per million (ppm) by weight of concrete. Throughout the remainder of this text, chloride ion concentration data are presented in units of ppm by weight of concrete Sample Point Locations - Beam 1 Chloride ion samples were collected at points along both sides of beam 1 as shown in Figure 43. The two crosshatched areas of the span indicate the possible regions of the girder directly above the highway surface depending upon the original orientation of the girder. Sides of the beam were labelled A and B. Figure 43 shows the beam as viewed from side A. Samples were taken on both sides of the. beam. The samples from side B were directly behind those on side A. Samples were referred to by beam, grid lin~, sample point (web or flange) and side. The points designated as number 3 were located in the web and points designated as number 2 were located in the flange. For example, a sample taken from 87

101 beam 1 at grid line 18, at mid-height of the web on side A was referred to as sample point l-18-3a. For beam 1, samples were taken from both sides at mid-height of the web and the bottom flange (points 3A, 3B, 2A, and 2B) at grid lines 16, 18, and 2. At grid lines 1, 13, 24 and 29, samples were taken from the bottom flange on both sides of the beam (points 2A and 2B). At points 2A and 2B on the bottom flange, samples were collected at 1/2 in., 1 in., 1-1/2 in., and 2 in. Concrete cover over prestressing strands at this location was 1-3/4" in. In the web, samples were also taken at 1/2 in. increments to a depth of 2-1/2 in. to 3 in. Concrete cover over draped prestressing strands in the web was 2 in Chloride Ion Concentrations - Beam 1 Chloride ion concentrations from the interior bridge girder (beam 1) are shown in Table 24 and Figures 44 through 5. Each figure shows chloride content versus sample depth for the all sample length of points drilled at the same grid line along the beam. The solid lines on the graphs represent samples obtained from the flange (points 2A and 2B) and the dotted lines, samples obtained from the web (points 3A and 3B). The lines are identified as to whether 88

102 they correspond to samples taken from side A or B of the beam. Only at the sample depth nearest the surface (1/2 in.) did chloride ion concentrations exceed levels of 3 ppm. The highest chloride ion concentration at the level of the steel (2 in.) was 27 ppm at point 1-2A. This value is suspect because for the same point at depths of 1 in. and 1-1/2 in. concentrations were 1 and 9 ppm, respectively. All other values at the level of steel were less than or equal to 12 ppm. At depths of 1/2 in. and 1 in. for all sample points, the chloride levels for side B were higher than levels for the corresponding point at side A. This seems to indicate that side B received higher exposure to deicing salts leading to the conclusion that it was most likely facing the incom:i,ng lane of traffic. Additionally, as might have been expected, at grid lines 16 and 18 where samples were taken at both the flange and the web, chloride ion concentrations were higher in the flange (2A and 2B) than in the web (3A and 3B) at both 1/2 in. and 1 in. depths. At grid line 2, chloride ion concentrations were significantly lower at 1/2 in. and 1 in. depths than anywhere else along the beam. It was determined by referring 89

103 to bridge layout drawings that this area was possibly covered by one of the intermediate diaphragms thus preventing its exposure to salts. It was not completely certain based upon the chloride levels, which section of the span was located directly over the roadway surface. However, in general, chloride levels at 1/2 in. depths for the flange were higher from grid lines 1 through 18 than for grid lines 2 through 29. Possibly this area was above the roadway surface and thus had greater exposure to the saltwater spray and mist carried forward by the traffic Sample Point Locations - Beam 2 Beam 2 was the facia girder facing incoming traffic. Sampl~ points on beam 2 were located on the web on either side and behind where the sign had been located and on both sides of the beam at intervals along the flange and the web. Figure 51 shows sample point locations for beam 2. The cross-hatched area shows the section of the span that was originally directly above the roadway. The beam is shown in Figure 51 as viewed from side A. Sample points from side B were located on the side of the beam not shown (side B) directly behind their depicted side A location on the 9

104 drawing. A labelling scheme similar to that used for beam 1 was employed to identify the sample points for beam 2. Sample points were referred to by beam designation (beam 2), a number giving horizontal location (see Figure 51), sample point (web = 3 or flange 2) and side (A or B). Because a grid system was not used for beam 2, horizontal locations were- designated by consecutive numbers from 1 through 9 ' distances are shown in Figure 51. Side B was the facia side of beam 2. Samples were collected at 1/2 in. intervals to a depth of 2 in. at all sample points Chloride Ion Concentrations - Beam 2 Chloride ion concentration data for girder 2 are given in Table 25 and Figures 52 through 57. Figures 52 through 56 show chloride ion concentration versus sample depth for sampl~ points at locations 2-6. The solid lines represent results for flange sample points (points 2A and 2B). The dotted lines represent results for web sample points (points 3A and 3B). At 1/2 in. and 1 in. depths, chloride levels for points on side B were greater or approximately equal to the chloride levels for corresponding points on side A. The only exception to this was at location 6, where the chloride ion concentration at the 1/2 in. depth for point 2-6-2B was lower than the other results along side B of the bottom 91

105 flange (28 ppm for point 2-6-2B versus 83 and 53 ppm for points 2-5-2B and 2-7-2B on either side). Excluding this case, the data generally supports the theory that the side of the beam facing incoming traffic gets higher exposure to salt spray than the other side of the beam. Except at locat-ion 6, at 1/2 in. depths, web chloride ion concentrations were lower than chloride ion concentrations for the flange. Figure 57 shows chloride ion concentrations for the points at locations 1, 7, 8 and 9. At each of these locations there was only one sample point drilled on side B (the facia or outer side) of the bottom flange. Chloride ion concentrations at 1/2 in. for these points ranged from 33 to 53 ppm. The higher values corresponded to the points nearer the area above the roadway surface and the lower values to the points further away from this area. Considering all the flange sample points, again, the locations which had higher chloride ion concentrations at a 1/2 in. depth were nearest the center of the section of the span originally over the roadway surface. From highest to lowest chloride ion concentrations, these point locations were in order: point 2-5-2B located 2.25 ft. from the center of the cross-hatched area (83 ppm), 92

106 point 2-4-2B located 4.75 ft. away (72 ppm), point 2-3-2B located ft. away (7 ppm)' point 2-2-2B located ft. away (6 ppm)' point 2-7-2B located ft. away (53 ppm)' point 2-8-2B located ft. away (49 ppm)' and point 2-9-2B located ft. away (33 ppm). Point 2-6-2A is not listed above because of the suspiciously low reading obtained at this point. Point 2-4-3B which was located in the web, behind where the sign had been located, had a chloride ion concentration of 21 ppm at 1/2 in. This was lower than the chloride ion concentrations for points 2-3-3B and 2-5-3B on the web on either side of the sign location (51 and 5 ppm, respectively). At 1/2 in. depths, point 2-6-3B on the web but further from the sign had a chloride ion concentration of 33 ppm and point 2-2-3B which was on the web but not over the roadway surface had a chloride ion concentration of 2 ppm. At depths of 1-1/2 in. and 2 in., chloride ion concentrations were less than or equal to 12 ppm for all sample points checked on beam 2 except for a high reading of 43 ppm for point 2-4-3A at 2 in. depth. This concentration significantly exceeded chloride levels for all other depths 93

107 at this same point including the 1/2 in. depth (21 ppm) and was therefore suspect. A1 though chloride ion concentrations at 1 in. depths and greater were comparable for beam 1 and beam 2, at 1/2 in. depths, chloride ion concentrations for beam 1 were significantly higher than for beam 2 in many cases. This difference was especially apparent when the portions of each beam with the highest salt exposure were compared with each other. The highest concentrations for beam 1 in the flange were in the range of 1 - were in the range of 7-12 ppm, in beam 2 these values 83 ppm. This result was opposite of the expected result. It seemed that an interior girder had greater salt exposure than an exterior girder facing incoming traffic. One possible explanation may be that exterior girders may have had salt washed off by rain in the spring and summer while this does not occur in the case of the interior girders Sample Point Locations - Beam 3 The locations of the seven sample points for girder 3 are shown in Figure 58. The facia side of Beam 3 is shown. All sample points were located on the facia side of the bottom flange. Cross-hatching shows two areas, one of which 94

108 was located over the roadway depending upon whether the girder was on the edge of the bridge over incoming or outgoing traffic. Points were simply numbered 1 through 7 and samples were taken at 1/2 in. increments to a depth of 2 in Chloride Ion Concentrations - Beam 3 Chloride ion concentration data for girder 3 are shown in Table 26 and Figure 59. In general, the chloride ion gradients obtained for all points along this beam were similar. The highest reading observed was 43 ppm at 1/2 in. depth at point 4 which was located at the span centerline. The chloride ion concentrations for beam 3 were lower than the values in the high exposure areas of either beam 1 or beam 2. Beam 3 concentrations were more comparable to the valuee; observed on side A (the low exposure side) of both beam 1 and beam 2. It was therefore likely that the facia side of beam 3 was facing toward outward flowing traffic and received less salt exposure than the other bridge girders tested. 95

109 Base Chloride Ion Concentrations Several chloride ion concentration samples were taken from a fourth bridge girder at the center of the beam section to determine the chloride ion concentration inherent in concrete itself not due to chloride ion penetration. The chloride ion concentrations in the web at depths of 3-4 in. for five sample points ranged from 3-5 ppm. (The web thickness was 7 in.) This base level was less than chloride ion concentrations measured at depths of 2 in. in beams 1, 2 and 3. Therefore, it was apparent that there was some penetration of chloride ions (although not a critical level see Tables 24, 25, 26) at the depth of the steel in the girders Chloride Concentrations in Bridge Deck The other girders from the bridge in this study that were not brought into the laboratory for testing were stored outdoors in a construction yard. Five sample points for checking chloride ion concentrations in the bridge deck were drilled in the sections that remained intact on several of these bridge girders. From the inspection reports of the bridge, it was determined that this was the original bridge deck dating from It was expected that chloride ion 96

110 concentrations for the bridge deck would be higher than chloride ion concentrations measured in the bridge girders for two reasons: 1) Salts were applied directly to the bridge deck and 2) the bridge deck concrete was of a lower strength and thus probably had greater permeability. Figure 6 shows the location of the sample points. Samples were drilled from the deck above three girders; two of the 64 ft. 8 in. facia girders and one 42 ft. interior girder. The beams are shown on this drawing as they were located in the yard. The deck above the short interior girder was located at the centerline of the bridge (a painted yellow roadway centerline was still visible on the deck above the girder). Sample point 3 was drilled from the deck above a long facia girder which is the same girder referred to as beam 3 in the preceding sections of this text. Slab thickness was a minimum of 6 in. for the 64 ft. 8 in. spans and a minimum of 7-3/4 in. for the 42 ft. spans. Minimum clearance to the top layer of bridge deck reinforcing steel was 1-1/2 in. Samples were drilled nearly above the centerline of the girders and were collected at depths of approximately 1/2 in. intervals to depths of about 3 in. Results from analysis of the samples are shown in Table 27 and Figure

111 Except for sample point 2, chloride ion levels for the bridge deck were much higher than for the bridge girders. Points 4 and 5 were located on the bridge deck centerline and these two points had the greatest amount of chloride ion penetration. At the level of first layer of bridge deck reinforcing steel (1-1/2 in.), chloride ion concentrations were -as high as 1-2 ppm by weight of concrete. There was some evidence of rust stains along the broken edges of the bridge deck. However no severe spa1ling or cracking of the bridge deck was visible Summary of Chloride Ion Tests Chloride ion penetration to 1/2 in. depths of all three bridge girders tested was observed to varying degrees depending upon the exposure of the concrete surface to deicing salts. At 1-1/2 in. depths and greater, chloride ion concentrations were not found to significantly exceed 25 ppm by weight of concrete, the lowest threshold level to initiate corrosion (see Section 5.1 of this text). The single exception to this was point 2-4-3B at 2 in. which contained 43 ppm chlorides. This value was considered to be suspiciously high based upon chloride ion concentrations at shallower depths for the same point. 98

112 The only evidence found of steel corrosion of the girders existed at the end faces where the epoxy coating over the beam ends had been chipped, thereby to exposing the ends of the prestressing strands to the elements. Rust stains were visible on the concrete at this location but spalling was not evidenced. As the girders were pretensioned and the strands were bonded to the concrete for the length of the beam, the prestressing strands were not exposed to the environment at any other locations. There was no other evidence of girder steel corrosion. Chloride ion penetration in the bridge deck was found to range from 111 to 194 ppm by weight of concrete at 1-1/2 in. depth, the depth of the first steel layer. The cement factor for the deck concrete was unknown. However, if a cement factor of 2% is assumed, chloride ion concentrations at this depth were ppm by weight of cement;. Although these levels exceeded chloride ion limits for bridge decks given in Section 5.1 (bridge decks to be replaced at greater than 3 ppm by weight of cement per the Federal Highway Administration [29]), no significant deterioration of the bridge deck was apparent from the portions of deck intact over the girder. 99

113 6. Conclusions Based upon the nondestructive and destructive test data, the following conclusions regarding the compressive strength of concrete in the bridge girder and the adequacy of the test methods to measure compressive strength can be made. 1) There was a 15 to 2 percent difference between the concrete compressive strength values obtained from the 4 in. diameter cores and the 2 in. diameter cores. This difference was observed after the strength values had been corrected for slenderness effects using the ASTM C42 correction factors. Core strengths were also adjusted to mid- depth of the girder to account for the variation in concrete strength with distance from the bottom of the concrete pour. Based on compression_ tests of 4 in. diameter cores drilled from the first bridge girder, concrete compressive strength averaged approximately 87 psi. For the 2 in. cores from this girder, concrete compressive strength averaged approximately 1,2 psi. This discrepancy was also observed for cores from a second identical girder. Cores from the second girder were drilled and tested by a second individual and resulted in 1

114 similar values. Average compressive strengths from the second girder were 842 and 9693 psi for the 4 in. cores and the 2 in. cores respectively. 2) Regression analysis was used to determine the relationship between measured strength and L/d ratio based on all data points (both 4 in. and 2 in. cores) from the two girders tested. A power curve model was used and the correlation coefficient, r, was. 866 (r 2.75). The correction factors obtained as a result of this analysis were less than both ASTM and BSI correction factors resulting in a greater correction for measured strength. 3) For the cores from the first girder, the 4 in. cores had been prepared for testing by capping the ends. The 2 in. cylinder ends were ground plane and parallel. In order to determine if the discrepancy in strength was due to differences in capping material, half of the six 2 in. cores from the second girder were capped and half were ground. The ground-end core values fell close to the L/d ratio versus measured strength curve-fit for all of the data points. The average measured strength value for the capped end cores fell close to this curve, 11

115 but the individual data points varied a great deal from the curve. (Measured strengths for the cappedend cores which all had L/d ratios of approximately 2 ranged from 643 to 9688 psi.) Because the number of capped end cores was small, a conclusion could not be reached concerning the affect of capped versus cored ends, however, it appeared that there was a greater variability in measured strength for the capped-end than for ground-end cores. 4) Based on results from the 2 in. core compress ion tests corrected per ASTM C42, the strength gain from 1 month to 2 years was 52 percent. This figure was higher than results obtained in past studies of concrete strength gain over time (27 to 37 percent strength gain for high early strength cement observed by Washa and Wendt over the period 1 month to 25 years [26]). Based on results from the 2 in. core compression tests corrected by the derived factors of this study, the strength gain over the time period in question was approximately 22 percent. 5) Results of linear regression analysis indicates a weak correlation between rebound number and concrete strength (r2=.26,.34 for strength 12

116 values corrected for L/d by ASTM C-42 and by the calculated factor, respectively). A better correlation was not obtained because of the relative uniformity of concrete strength in the girder. The rebound hammer manufacturer's equation underestimated concrete strength by an average 22 percent based on ASTM C42-corrected strength. For average values, the manufacturer's relationship was only 2.6 percent different from the core strength corrected L/d correction factors derived in this text. 6) The sensitivity of the pulse velocity equipment used was not sufficient to distinguish compressive strength over the range of strengths measured. The theoretical equation with a K-value based on the dynamic Poisson's ratio of concrete predicted concrete strength on average within 9 percent for ASTM C-42 corrected strength. This was better than Facaoaru's equation for the reference concrete which greatly underestimated concrete strength based on this value. However, Facaoaru's relationship was calibrated with the calculated C t factor to obtain a curve that matched the behavior of the concrete tested. The theoretically derived pulse velocity strength relationship predicted an average 13

117 strength 36 percent greater than the average strength obtained when the L/d correction factors from Equation 2 were used. Facaoaru's relationship based on reference concrete underestimated strength for calculated factor-corrected strength. 7) No relationship between exposed probe length of the Windsor probe method and concrete strength was found in the range of concrete strength measured. This was attributed to the relative uniformity of concrete tested. The manufacturer's equations greatly underestimated concrete strength (corrected by both methods of this text). 8) No conclusions could be drawn from break-off test data concerning concrete strength due to the great variability of test data. The literature states that this test cannot be used for concrete made with greater than 3/4 in. aggregate. The aggregate in the girder concrete mix under study was 1 in. maximum size. This fact as well as the suspicion that some damage may have occurred during drilling of the break-off cores accounts for the fact that results from break-off tests were much less than expected based on the manufacturer's curves. Problems in drilling were probably due at least 14

118 partly to the hardness of the concrete and probably would not have occurred if testing a weaker concrete. 9) The coefficients of variations for the tests performed were relatively low indicating fairly uniform concrete compressive strength. No particular area of weak concrete was identified based on the tests and there did not seem to be a difference in strength of concrete between the cracked and uncracked sections of the beam. 1) Of the tests, the rebound hammer and pulse velocity tests predicted concrete strength reasonably well on average based upon existing correlations. Manufacturer's relationships for the Windsor probe were not able to predict the strength. The break-off test results were found to be invalid due to the large concrete aggregate size and damage to the break-off cores during the drilling process. ll) Varying chloride ion concentration levels were found at 1/2 in. depths depending on the exposure of the section of the bridge girder to deicing salts. At shallow depths, chloride ion concentrations in the interior girder tested (beam 1) exceeded those in the facia girder facing incoming traffic (beam 2). This was attributed to rain having washed of salts from the exterior face of the beam and therefore reducing the period of time that the concrete was in contact with salt during each year. 15

119 12) At the level of the reinforcement, the majority of the chloride ion measurements indicated concentrations less than 1 ppm. The highest chloride concentrations measured were not found to significantly exceed the corrosion threshold concentration (25 ppm by weight of concrete) with one exception. The exception, point 2-4-3B with 43 ppm chlorides at a depth of 2 in., was considered suspicious. Chloride ion concentrations measured at shallower depths for the same point were consistently lower (i.e. 21 ppm at 1/2 in. depth, 2 ppm at 1 in. depth, and 7 ppm at 1-1/2 in. depth). 13) The only evidence of steel corrosion observed in the girders was at the end faces where the epoxy coating over the beam ends had been chipped to expose the ends of the prestressing strands. Rust stains were visible on the concrete at this location, but spalling was not observed. 14} Chloride ion concentrations of the bridge deck were found to be elevated well above the chloride ion concentration threshold for corrosion at the level of the steel. Nevertheless, obvious visual evidence of corrosion was not found. In general, it was confirmed that the nondestructive tests of this study must be calibrated in order to yield accurate concrete strengths and that these tests are best used as an indication of concrete uniformity. Further work on the break-off tester is needed to further delineate its 16

120 limitations for testing high strength and/or large aggregate concretes. Core testing gave different results for 2 in. and 4 in. cores which led to the derivation of L/d correction factors that differed from ASTM C42 factors. Whether this was due to core size only or core size/aggregate ratio was not apparent. Results of chloride ion testing showed that chloride ion penetration into the bridge girders was much less than for the bridge deck. In the girders, chloride ion concentrations measured at the reinforcement level had not reached a critical level after 2 years of service. 17

121 REFERENCES 1. Greene, Gordon W., "Test Hammer Provides New Method of Evaluating Hardened Concrete," ACI Journal, Vol. 51, Nov. 1954, pp Malhotra, V.M., Testing Hardened Concrete: Nondestructive Methods (ACI Monograph No. 9), American Concrete Institute, American Society for Testing and Materials, ASTM C85, "Standard Test Method for Rebound Number of Hardened Concrete." 4. Zoldners, N.G., "Calibration and Use of Impact Test Hammer," Proceedings, American Concrete Institute, Vol. 54, August, 1957, pp Samarin, A. and Dhir, R.K., "Determination of In Situ Concrete Strength: Rapidly and Confidently by Nondestructive Testing," In Situ/Nondestructive Testing of Concrete, V.M. Malhotra, ed., ACI Publication SP-82, 1984, pp Facaoaru, I., "Romanian Achievements in Nondestructive Strength Testing of Concrete," In Situ/Nondestructive Testing of Concrete, V.M. Malhotra, ed., ACI Publication SP-82, 1984, pp American Society for Testing and Materials, ASTM C597, "Standard Test Method for Pulse Velocity Through Concrete." 18

122 8. Whitehurst, E.A., Evaluation of Concrete Properties from Sonic Tests (ACI Monograph No. 2), American Concrete Institute, Gardiner, R.A. and Hatcher, D.S., "Material and Dimensional Properties of an Eleven-Story Reinforced Concrete Building," Washington University, Department of Civil Engineering, Research Report No. 52, Structural Division, August, Tanigawa, Y., Baba, K., and Mori, H., "Estimation of Concrete Strength by Combined Nondestructive Testing Method," In Situ/Nondestructive Testing of Concrete, V.M. Malhotra, ed., ACI Publication SP-82, 1984, pp Nawy, E.G. and Balaguru, P.N., "High-strength Concrete," Handbook of Structural Concrete, F.K. Kong, R.H. Evans, E. Cohen and F. Roll, ed., McGraw-Hill, New York, Pauw, A., "Static Modulus of Elasticity of Concrete as Affected by Density," Journal of the American Concrete Institute, Vol. 32, No. 6, December, 196, pp Sturrup, V.R., Vecchio, F.J. and Caratin, H., "Pulse Velocity as a Measure of Concrete Compressive Strength," In Situ/Nondestructive Testing of Concrete, V.M. Malhotra, ed., ACI Publication SP-82, 1984, pp Mikhailovsky, L. and Scanlon, A., "Evaluation of Existing Bridge Structure by Nondestructive Test Methods," Structural Engineering Report No. 128, Depart- 19

123 ment of Civil Engineering, The University of Alberta, American Society for Testing and Materials, ASTM C83, "Standard Test Method for Penetration Resistance of Hardened Concrete." 16. Scancem Chemicals, "Break-off Tester User's Guide," A/S gcancem Chemicals, N-347 Slemmestad, Norway. 17. Dahl-Jorgensen, E. and Johansen, R., "General and Specialized Use of the Break-off Concrete Strength Testing Method," In Situ/Nondestructive Testing of Concrete, V.M. Malhotra, ed., ACI Publication SP-8.2, 1984, pp Barker, M.G. and Ramirez, J.A., "Determination of Concrete Strengths Using the Break-off Tester," Report CE-STR-87-22, Purdue University School of Civil Engineering. 19. Munday, J. and Dhir, R., "Assessment of In Situ Concrete Quality by Core Testing," In Situ/Nondestructive Testing of Concrete, V.M. Malhotra, ed., ACI Publication SP-82, 1984, pp American Society for Testing and Materials, ASTM C42, "Standard Method of Obtaining and Testing Drilled Cores and Sawed Beams of Concrete." 21. British Standards Institution, B.S. 1881: Part 12, Method for determination of the compressive strength of concrete cores. BSI, London,

124 22. Kesler, C.E., "Effect of Length to Diameter Ratio on Compressive Strength," ASTM Bulletin, No. 221, April, 1957, pp Bloem, D.L., "Concrete Strength in Structures," Journal of the American Concrete Institute, Vol. 65, March 1968, pp.l Swamy, R.N. and Al-Hamed, A.H., "Evaluation of Small Diameter Core Tests to Determine In Situ Strength of Concrete," In Situ/Nondestructive Testing of Concrete, V.M. Malhotra, ed., ACI Publication SP-82, 1984, pp Werner, G., "The Effect of Type of Capping Material on the Compressive Strength of Concrete," Proceedings of the American Society of Testing and Materials, Vol. 58, 1958, pp Soroka, I., Portland Cement Paste and Concrete, Chemical Publishing Co. Inc., 198, 338 pp. 27. W~sha, G.W. and Wendt, K.F., "Fifty Year Properties of Concrete," Journal of the American Concrete Institute, Vol. 72, January 1975, pp Slater, J.E., Corrosion of Metal in Association with Concrete, ASTM Special Technical Publication 818, Locke, C.E., "Corrosion of Steel in Portland Cement Concrete: Fundamental Studies," ASTM STP 96, Corrosion of Rebars in Concrete, V. Chaker, ed.,

125 3. Berman, H.A., "The Effect of Sodium Chloride on the Corrosion of Concrete Reinforcing Steel and on the ph of Calcium Hydroxide Solution," FHWA-RD-74-1, Federal Highway Administration, Washington, DC, Hausmann, D.A., "Steel Corrosion in Concrete: How Does It Occur," Materials Protection, November 1976, p Glear, K.C., "Time to Corrosion of Reinforcing Steel in Concrete Slabs," FHWA-RD-76-7, Federal Highway Administration, Washington, DC, Spellman, D.L., and Stratfull, R.F., "Concrete Variables and Corrosion Testing," Highway Research Record, No. 423, 1973, p Pfeifer, D.W., Langren, J.R. and Zoob, A., "Protective Systems for New Prestressed and Substructure Concrete," FHWA-RD , Federal Highway Administration, Mclean VA, VandeVeer, J.R., "Techniques for Evaluating Reinforced Concrete Bridge Decks." Journal of the American Concrete Institute, Vol. 63, 1966, p Browne, R.D., Geoghegan, M.P. and Baker, A. F., "Analysis of Structural Condition from Durability Results," Corrosion of Reinforcement in Concrete Construction, Alan P. Crane, ed., Ellis Horwood Limited, Chichester, England, 1983, pp.l Allen, R.T.L., and Forrester~ J.A., "The Investigation and Repair of Damaged Reinforced Concrete Structures," 112

126 Corrosion of Reinforcement in Concrete Construction, Alan P. Crane, ed., Ellis Horwood Limited, Chichester, England, 1983, pp Griess, J.C. and Naus, D.J., "Corrosion of Steel Tendons Used in Prestressed Concrete Pressure Vessels," Corrosion of Reinforcing Steel in Concrete, ASTM Special Technical Publication 713, 198, pp Hodgkiess, T., "Fatigue of Reinforced Concrete," Corrosion of Steel Reinforcements in Concrete Construetion, Society of Chemical Industry, London, 1979, pp Berman, H.A. "Determination of Chloride in Hardened Portland Cement Paste, Mortar and Concrete," Federal Highway Administration, Interim Report, FHWA-RD Walpole, R.E. and Myers, R.H., Probability and Statistics for Engineers and Scientists, Macmillan Publishing Co., Inc., New York,

127 TABLES

128 TABLE 1 WINDSOR PROBE MANUFACTURER'S COMPRESSIVE STRENGTH RELATIONSHIP COMPRESSIVE STRENGTH (PSI) EXPOSED PROBE MOHS' MOHS' MOHS' MOHS' MOHS' (IN.) NO. 3 NO. 4 NO. 5 NO. 6 NO

129 TABLE 2 REBOUND HAMMER DATA SERIES I - BOTTOM SURFACE OF BOTTOM FLANGE BEFORE BEAM FATIGUE TESTING GRID COEFFICIENT LOCATION AVERAGE MINIMUM MAXIMUM VARIATION OF * * * * *One value thrown out according to reject criteria 115

130 TABLE 3 REBOUND HAMMER DATA SERIES II - SIDE SURFACE OF BOTTOM FLANGE BEFORE BEAM FATIGUE TESTING GRID LOCATION *A--1 B-2-3 A-4-5 *B-6-7 A-8-9 B-1-11 A B A *B A-2-21 B A B *A B-3-31 COEFFICIENT of AVERAGE MINIMUM MAXIMUM VARIATION (%) * One value thrown out according to reject criteria 116

131 TABLE 4 REBOUND HAMMER DATA SERIES III - SIDE SURFACE OF TOP FLANGE BEFORE BEAM FATIGUE TESTING GRID COEFFICIENT of LOCATION AVERAGE MINIMUM MAXIMUM VARIATION (%) A B A B A *B A *B *A B A *B *A **B **A B * One value thrown out according to reject criteria ** Two values thrown out according to reject criteria 117

132 TABLE 5 REBOUND HAMMER DATA SERIES IV - WEB BEFORE BEAM FATIGUE TESTING GRID LOCATION *A-1-2 B-3-4 *A-5-6 B-7-8 A-9-1 *B A *B **A B-19-2 A B A B A-29-3 AVERAGE MINIMUM MAXIMUM COEFFICIENT of VARIATION (%) * One value thrown out according to reject criteria ** Two values thrown out according to reject criteria 118

133 TABLE 6 REBOUND HAMMER DATA BEFORE BEAM FATIGUE TESTING SUMMARY TEST COEFFICIENT of SERIES AVERAGE MAXIMUM MINIMUM VARIATION (%) *I II III IV *READINGS TAKEN IN VERTICAL DIRECTION. TAKEN HORIZONTALLY. ALL OTHER READINGS 119

134 TABLE 7 REBOUND HAMMER DATA AT CORE LOCATIONS AFTER BEAM FATIGUE TESTING COEFFICIENT of CORE AVERAGE MINIMUM MAXIMUM VARIATION (%) * *5-l * * * * ** Only 5 readings taken. Side B of beam inaccessible and three points on side A located near Windsor probe spall area. 9 Readings only. Side B of beam inaccessible. * One value thrown out according to reject criteria ** Two values thrown out according to reject criteria AVERAGE REBOUND NO. COV OF REBOUND NO. MAXIMUM REBOUND NO. MINIMUM REBOUND NO % AVG. STRENGTH= 796 PSI COV OF STRENGTH= 4.17% STRENGTH VALUE BASED UPON MANUFACTURER'S CORRELATION EQUATION. NOTE: CORE 12-1 WAS NOT DRILLED. 12

135 TABLE 8 WINDSOR PROBE DATA TEST EXPOSED PROBE LENGTH (IN) AREA RANGE AVERAGE * * ** *** ** *.9 2.2** * Sample point not obtained due to space constraints. ** Average of two values. *** Value discarded. AVERAGE READING = 2.19 IN. COV OF READINGS = 3.6% AVERAGE STRENGTH BASED UPON MANUFACTURER'S EQUATION FOR A MOH'S HARDNESS OF 6 (MOH'S HARDNESS 7) = 695 PSI (6395 PSI) COV OF STRENGTH BASED UPON A MOH'S HARDNESS OF 6 (MOH'S HARDNESS 7) = 9.15% (1.81%) 121

136 TABLE 9 PULSE VELOCITY DATA BEFORE AND AFTER CORING (PULSE VELOCITIES in THOUSAND ft/s) CORE vcorel vcore vin situ vin situ AVERAGE cov STD. DEV % 1. 33% AVG. STRENGTH 11 PSI 17 PSI COV STRENGTH 5.51% 5.28% NOTE: STRENGTH CALCULATIONS BASED UPON EQN. 9, K=

137 TABLE 1 BREAK-OFF TEST DATA AREA AVG COV(%) f' (psi) llo llo AVERAGE OF 12 SETS 87.5 COV OF 12 AVERAGES 12.6% AVERAGE STRENGTH = 374 PSI COV OF STRENGTH = 21.5% * Based on manufacturer's break-off reading/strength relationship (f' c = 1.4 x B1. 76 ) 123

138 TABLE 11 L/d RATIO CORRECTION FACTORS LENGTH/ DIAMETER ASTM BSI RATIO FACTOR FACTOR

139 TABLE 12 4 IN. CORE STRENGTH CORRECTED TO GIRDER MID-DEPTH 4 IN. CORE MEASURED IN. FROM CORRECTION STRENGTH BOTTOM FACTOR FOR (PSI) OF WEB DEPTH CORRECTED STRENGTH (PSI)

140 TABLE 13 4 IN. CORE COMPRESSION TEST RESULTS DEPTH CORR. CORRECTED AVERAGE AVERAGE STRENGTH CORRECTION STRENGTH CORE LENGTH DIAMETER L/d (PSI) FACTOR (PSI) l * * Sample not tested. 126

141 TABLE 14 2 IN. CORE MEASURED STRENGTH CORRECTED TO GIRDER MID-DEPTH BREAK- OFF CORE MEASURED IN. FROM CORRECTION STRENGTH BOTTOM FACTOR FOR (PSI) OF WEB DEPTH DEPTH CORRECTED STRENGTH

142 TABLE 15 2 IN. CORE COMPRESSION TEST RESULTS BREAK- DEPTH L/d OFF AVG. AVG. CORRECTED CORRECTION CORE LENGTH DIAMETER L/d STRENGTH FACTOR* CORRECTED STRENGTH (PSI) : AVERAGE CORRECTED STRENGTH = 1216 COEFFICIENT OF VARIATION = 8.54% *ASTM C42 correction factors (.87 for L/d = tq obtain factors for L/d other than 1). 1; interpolating 128

143 TABLE 16 4 IN. CORE COMPRESSION TEST RESULTS - BEAM 2 MEASURED DEPTH ASTM CORRECTED STRENGTH CORRECTED L/d CORRECTION STRENGTH CORE (PSI) STRENGTH RATIO FACTOR (PSI) 4-1* * * * Capped ends AVERAGE CORRECTED STRENGTH = 843 PSI STANDARD DEVIATION 1255 PSI COEFFICIENT OF VARIATION 15.6% 129

144 TABLE 17 2 IN. CORE COMPRESSION TEST RESULTS - BEAM 2 MEASURED DEPTH STRENGTH CORRECTED L/d CORE (PSI) STRENGTH RATIO ASTM CORRECTION FACTOR CORRECTED STRENGTH (PSI) AVERAGE CORRECTED STRENGTH = 9693 PSI STANDARD DEVIATION = 841 PSI COEFFICIENT OF VARIATION = 8. 7% 13

145 TABLE 18 REGRESSION ANALYSIS AVERAGE REBOUND NUMBER VERSUS COMPRESSIVE STRENGTH TEST AVERAGE 2 IN. CORE AREA REBOUND # STRENGTH(PSI)* R f'c RESULTS OF REGRESSION** Y (PSI) *CORRECTED TO DEPTH OF REBOUND READINGS ** Y = R- 377 ; r 2 =.26 MEAN ST[l. DEV. cov % % 131

146 TABLE 19 COMPARISON OF REBOUND NUMBER - STRENGTH RELATIONSHIPS ASTM C42 L/d CORRECTION FACTORS % DIFFERENCE BETWEEN CORE STRENGTH AND EQNS. IN PARENTHESES TEST AVERAGE 2 IN. CORE f'c FROM f'c FROM -AREA REBOUND # ASTM-CORR. MANUFACTURER FACAOARU R f'c (PSI) EQN. EQN (27%) 913 (15%) (22%) 9488 (8%) (24%) 115 (9%) (18%) 1114 (.4%) (18%) 8426 (7%) (25%) 9818 (1%) (29%) 8754 (18%) (13%) 988 ( -1%) AVG (22%) 9315 (9%) 132

147 TABLE 2 COMPARISON OF REBOUND NUMBER - STRENGTH RELATIONSHIPS EQUATION 2 L/d CORRECTION FACTORS % DIFFERENCE BETWEEN CORE STRENGTH AND EQNS. IN PARENTHESES TEST AVERAGE 2 IN. CORE f'c FROM f'c FROM -AREA REBOUND # EQN.2-CORR. MANUFACTURER FACAOARU R f'c (PSI) EQN. EQN (7%) 913 (-8%) (4%) 9488(-13%) (3%) 115(-17%) (-2%) 1114(-24%) (-2%) 8426(-15%) (4%) 9818(-15%) (9%) 8754 (-4%) (-7%) 988(-24%) AVG (3%) 9315(-15%) 133

148 TABLE 21 PULSE VELOCITY VERSUS 2 IN. CORE STRENGTH AVG CORE 2 IN. CORE f' c FROM f'c FROM TEST PULSE VEL. STRENGTH EQN.9 EQN. 1 AREA (1 FT/S) (PSI) K =.85 REF CONC MEAN STD. DEV cov.91% 7.6% 3.6% 4. 6% 134

149 TABLE 22 WINDSOR PROBE READINGS VERSUS 2 IN. CORE STRENGTH EXPOSED 2 IN. CORE f'c FROM f'c FROM TEST PROBE LENGTH STRENGTH MANUF. EQN. MANUF. EQN. AREA (IN.) (PSI)* MOH'S #6 MOH'S # / / / / / / / MEAN / STD DEV / cov 4.3% 7.4%/6% 11% 13% *First value shown is strength corrected as per ASTM C42. Second value is strength corrected as per Equation

150 TABLE 23 BREAK-OFF READINGS VERSUS 2 IN. CORE STRENGTH BREAK-OFF CORE f'c CORE f'c f'c FROM CORE READING ASTM-CORR. SEC MANUF. EQN

151 TABLE 24 CHLORIDE ION CONCENTRATIONS (PPM)* GIRDER 1 TEST POINT 1-2A 1-2B 13-2A 13-2B 16-2A 16-2B 16-3A 16-3B 18-2A 18-2B 18-3A 18-3B 2-2A 2-2B 2-3A 2-3B 24-2A 24-2B 29-2A 29-2B DEPTH 1/2" 1" 1-1/2" 2" 2-1/2" " 5 NOTE: THE WEB WAS POINT 3 AND THE FLANGE WAS POINT 2 * BY WEIGHT OF CONCRETE 137

152 TABLE 25 CHLORIDE ION CONCENTRATIONS (PPM)* GIRDER 2 TEST POINT 1-2B 2-2A 2-2B 2-3B 3-2A 3-2B 3-3A 3-3B 4-2A 4-2B 4-3A 4-3B 5-2A 5-2B 5-3A 5-3B 6-2A 6-2B 6-3B 7-2B 8-2B 9-2B DEPTH 1/2" 1" 1-1/2" no " NOTE: SIDE B WAS THE SIDE OF THE BEAM FACING INCOMING TRAFFIC. THE WEB WAS POINT 3 AND THE FLANGE WAS POINT 2. * BY WEIGHT OF CONCRETE 138

153 TABLE 26 CHLORIDE ION CONCENTRATIONS (PPM)* GIRDER 3 TEST POINT DEPTH 1/2" 1" 1-1/2" " NOTE: ALL SAMPLE POINTS WERE ON THE FACIA SIDE OF THE GIRDER * BY WEIGHT OF CONCRETE 139

154 TABLE 27 CHLORIDE ION CONCENTRATIONS (PPM)* BRIDGE DECK PT. 1 IN. PPM PT. 2 PT. 3 PT. 4 IN. PPM IN. PPM IN. PPM PT. 5 IN. PPM 1/ / / / / /2 24 3/ / / / / / / / / / / / / /4 44 1/ / / / * BY WEIGHT OF CONCRETE 14

155 FIGURES

156 1-' ~ w ~ - 7 en 5 6 u 5 - ' f' c = E -1 3 K 2 V Jl I I I I I I I I I I I / I I I / / / / / / / / / / / / K=.9 I I K=.85 I I I f' = a e.335v c / / a=34.8 // // // / PULSE VELOCITY (ft/s) Fig. 3 Pulse Velocity/Strength Relationships

157 N d/v. 3,2 3,6 1,., lt;2, ti;6!.;;& "i km/.s Fig. 4 Combined Rebound Number and Pulse Velocity/Strength Relationships 144

158 Gage Top Plate Gage Base Plate Single Probe Measuring Cap Base Plate Single Probe Locating Template Ill Gage Base Plate Retainers Sore Brush~ Cleaning Rod Probe Locating Template Micrometer Gage Test Area Cleaning Brush Probe Removal Kit Driver Heads Fig. 5 Windsor Probe Apparatus 145

159 Fig. 6 Break-off Test Set Up 146

160 147

161 9 8 (f) ~7 u y_ 6 f'c x I f- 5 C) z ~ 4 f C/) w > 3 ~ 2 w :::: Q_ 1 2 u,_--,--,---,--,--,---,--,---,--,--,---,--, BREAK-OFF VALUE (BAR) Fig. 8 Break-off Value/Strength Relationships 148

162 "TRUE" CHARACTERISTIC STRENGTH , n>3, lit= % n= 5, 11t=9 /o n= 3,11t=3 /o I I I c. NORMAL VARI ATION IN-SITU "" COEFF. OF VARIATION OF CONCRETE lie (%) Fig. 9 Influence of Within Test Variation and Number of Tests on Calculated Strength 149

163 j----41'8.5' Gl,. 64'1' 64'1' 41'8.5' ~ ~ ~ I G7 2 Gl2 3 1 Gl3 G2 G9 Gl4 G5 GIS... V1 66 7' 1 GIRDERS LABELLED G1-G16 N PUN VID Of BRIDGE BOT TO SWI Fig. 1 Bridge Superstructure

164 7". j_ DRAPED STRANDS Ill 4 ~ 19// 7" i ~ Ell Ill i 4 1/2" l'- ld':_i STRAIGHT 7" STRANDS Fig. 11 Girder Cross-sectional Dimensions 151

165 DRAPED STRANDS ~R~PE~F sm:~~ -<::-:... ""' I I I 1-' U1.tv ~ ~ I STRAIGHT STRANDS POINT I IN~ 5' ( ± 1 I ) C END VIEW SIDE VIEW Fig. 12 Prestressing Strand Location

166 !:2 P:l- ::!; -- I ~- i ~-- Rl- ~ ~ T CD C':'>J I -!I@ tjl!; t:: - om ~-- -4 Pel- ~ ~ 4J <I en en en "' ~ I =-- -- ~- ~~...:l ~ 4J in?-.! Ill ~-- 1 II Q) I c::n-- 4 ' ~- :... :... ~. ~ "d I:G- t:: I ~ ~ =-- ~ r--- I ~- - "' - en ::J ~-- ;;::;- ~...:l "' I ~- ; "d ~ om... C!l ~-- ::>- =-- i!l ~ a.n-- I ~-- I 1:""1-- t">> ~ s~ ~ 85!::::::: _i_ 1--l!l~ ~.. <I m c:;;;-- Q) fia ~~~~Ill eno:nt;t; -~~;: ~ii;; ~ ~--... i i!i ~--... <I <I I ~--!:2 1--l!l~~ 4J :>-.. a "' j:q C"'"l.-I b()..-l!l. 153

167 Rebound Hammer Test Locations... l.11 ~ Slab -1'- 4" ,.. I t'- w 1 """... Sides of Beam Denoted A & B t 7 112' Grid Lines A I2'222J B I?22ZI --Test Series (II) I I I I llrld Blocks AH 81-J, / Test Series (Ill) Grid Blocks Al-2, 83-4, AS Test Series (IV) Grid Blocks Al , AS-6.. A '- 4' ' = 6'- '... 2'- 4'... M-S... BJG-31 I I I I '. : : ~. ----Test Series (I) 1 ~ Grid Blocks JG-31 Bottom Flange Note: Letters A and B indicate side of beam for each test group Fig. 14 Rebound Hammer Test Locations

168 7 112' 8 112' To Tap ll'f V<b To EmJe g : : 1 To Grid Line 'I 2 112:1 To Grid Line Test Series IV Test Series I 1' 2' 2' 1112' ~ @ ----''-.,..9112',j2".j,2',j To Grid Line Test Series II. III Fig. 15 Rebound Hammer Test Patterns 155

169 TEST SERIES BOTTOM FLANGE (UPWARD) 7 6 t-' Ul \ 5 * 4 z ::J ill 3 w II AVERAGE FT FROM END -t-- MINIMJM MAXIMJM Fig. 16 Rebound Hammer Test Results (Series I)

170 TEST SERIES II BOTTOM FLANGE (SIDE) 7 6 * z :::J [J) w t-' ([ ll1 " I AVERAGE FT FROM END --t- MINIMUM ~MAXIMUM Fig. 17 Rebound Hammer Test Results (Series II)

171 TEST SERIES Ill TOP FLANGE (SIDE) 7 6 :t-1: 5 4 z ::J m 3 w t-' :: Ul 2 1 I I I I I io FT FROM END -8- AVERAGE --+- MINIMUM ~MAXIMUM Fig. 18 Rebound Hammer Test Results (Series III)

172 TEST SERIES IV WEB 7 6 t-' U1 1.. it: 5 4 z ':J m 3 w a:: 2 io io FT FROM END AVERAGE --+- MINIMUM ~MAXIMUM Fig. 19 Rebound Hammer Test Results (Series IV)

173 H H H to H H = > C/1(/")(/") www * <l..q- z w 2 cr::: LL f Oll_ N N to..q- N # ONn83d 16

174 Rebound Hammer Tests at 4" Core Locations Note: Test pattern same on both sides of beam Fig. 21 Rebound Hammer Test Pattern for 4 in. Core Locations 161

175 6 I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I FT FROM END Fig. 22 Rebound Hammer Test Results at Core Locations after Fatigue Testing

176 6 :::ft::: I-' \ w 4 z ~ m w Cl:: 2!:;. BEFORE TESTING * AFTER TESTING I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I I FT FROM END Fig. 23 Comparison of Average Rebound Numbers before and after Fatigue Testing

177 164

178 HISTOGRAM OF PULSE VELOCITY.READINGS BEFORE CORING 7 ~ ~ 6 1-' \ V1 u > z w :::J w a:: u_ PULSE VELOCITY (1 FT/S) Fig. 25 Histogram of Pulse Velocity through Cores before Drilling

179 HISTOGRAM OF PULSE VELOCITY.READINGS AFTER CORING 7~ ~ 6 t-' ~ ~ > u z w ::J w a: LL PULSE VELOCITY ( 1 FT /S) Fig. 26 Histogram of Pulse Velocity through Cores after Drilling

180 Fig. 2 7 Drill Bit for Break-off Tests 167

181 Fig. 28 Drilling Apparatus Fig. 29 Drilling Apparatus 168