Ductility of Welded Steel Column to Cap Beam Connections Phase 2

Size: px

Start display at page:

Download "Ductility of Welded Steel Column to Cap Beam Connections Phase 2"

Emory McLaughlin
6 years ago
Views:

1 Ductility of Welded Steel Column to Cap Beam Connections Phase 2 Test 2 Summary Report 1 June Executive Summary Prepared for: Alaska Department of Transportation and Public Facilities Alaska University Transportation Center UAF Prepared by: Steven J. Fulmer Mervyn J. Kowalsky James M. Nau North Carolina State University, Raleigh, NC The test 2 specimen consisted of a composite connection configuration utilizing shear stud connectors placed on the outside the pile section which was inserted into a larger stub pile section that contained a matching pattern of shear studs as shown in Figure 4. The annular space formed between the two components was filled with flowable grout allowing for composite action to develop the moment connection by transferring forces from the pile to the cap beam via the stub pile, while avoiding inelasticity in critical welded regions. The results of experimental testing showed the connection capable of successfully developing a flexural hinging mode of failure in the form of pile wall local buckling at the intended region. All regions which were considered to be in critical need of remaining elastic were shown through data analysis to remain within the intended elastic range as the full flexural hinging capacity of the piles developed. Ultimately, the limit state of pile wall local buckling was shown to control at a reliable displacement ductility level of 3 associated with 8.44 of displacement or 8.1% drift. Although associated with large levels of damage and loss of strength, the specimen was able to remain in-tact at a displacement ductility of 6 associated with of displacement. 1

2. Introduction The second test of the second phase of this project was aimed at evaluating the capability of a composite connection between the pile and cap beam elements of the steel pier, to

2 2. Introduction The second test of the second phase of this project was aimed at evaluating the capability of a composite connection between the pile and cap beam elements of the steel pier, to effectively isolate damage in the form of flexural hinging within the pile section while protecting connection and cap beam elements as the pier was subjected to lateral loading. More specifically, the connection consisted of a 24 x0.500 stub pipe pile section that was connected to the cap beam by a complete joint penetration weld (CJP) with a 3/8 reinforcing fillet weld. The inner diameter of the stub pile contained 12 lines of welded 3/4 diameter 2-1/2 long shear stud connectors located at 30 degrees on centers around the pipe with 4 shear studs in a given line. Similarly the top of the HSS16x0.500 pile section had 12 lines of 4 matching shear studs welded at 30 degrees on centers around the cross section offset by 15 radially and 2-1/2 vertically from the studs inside the larger stub pile. The shear stud pocket, shown in Figure 1, was grouted utilizing non-shrinking flowable grout to complete the moment resisting connections of the bent as shown in Figure 2. Further design and construction details are provided in subsequent sections of this report. Figure 1. Annular Shear Stud Pocket 2

3 Figure 2. Steel Bent with Composite Connections 3. Bent and Connection Design Similar to past configurations, ASTM A500 Gr. B HSS16x0.500 piles were chosen as the column elements of the bent to produce reasonable aspect and D/t ratios. The construction of the cap beam consisted of double ASTM A572 HP14x117 sections to provide both a capacity protected cap beam as well as adequate bearing seat width for single span girders, should a designer choose not to utilize continuous spans. A 24 x0.500 pipe section manufactured to the material standards of ASTM A500 Gr. B was utilized as the stub pile element to provide an adequate gap for the placement of shear studs as well as to accommodate construction tolerances. The design of the composite connection was based on the assumption that the total nominal strength of the shear stud connectors, on the pile side, should be capable of developing yielding of the HSS16x0.500 gross cross section. From known, or anticipated, material properties a required number of 3/4 diameter shear studs could be determined and distributed around the cross section in an even pattern. A matching number of studs were then placed on the stub pile side to facilitate load transfer in a truss mechanism between the studs on either side of the connection. The nominal capacity of a single shear stud was determined utilizing the provisions of AASHTO LRFD Bridge Design Specifications (2007) Section as well as the AISC 13 th Edition Specifications Section I3 both of which provide the same model shown in Eq. 1. As indicated in 3

Eq. 1, the model is a function of concrete compressive strength as well as the cross sectional area of the shear stud with an upper bound of stud shear rupture.

4 Eq. 1, the model is a function of concrete compressive strength as well as the cross sectional area of the shear stud with an upper bound of stud shear rupture. Although the model is intended for use (in both codes) for composite construction between beams and a slab or bridge deck, it has been assumed conservative for use in this design given the highly confined nature of the annular grout pocket maintaining a logical upper bound of shear rupture of the stud.. Q n = 0.5A sc (f c E c ) 0.5 A sc F u E c =1746(f c ) 0.5 AISC, 1820(f c ) 0.5 AASHTO Eq. 1a (ksi) Eq. 1b (NWC) Utilizing this model along with expected material properties of the grout and specified material properties of ASTM A108 shear stud connectors, it was found that a minimum of 43 shear stud connectors were necessary based on an anticipated yield stress of 54 ksi for the HSS16x0.500 ASTM A500 Gr. B piles. To generate a symmetrical condition, 12 lines of 4 shear studs at 30 on centers were utilized providing 48 shear studs on the pile side or 96 total per connection as shown in Figure 3. The details of the connection are provided in Figure 4. It should also be noted that the 24 x0.500 stub pile section, acting non-compositely at the cap beam connection, was designed to remain elastic at the full flexural over strength capacity of the piles. Figure 3. Pile Side Shear Studs (48 per pile) 4

5 Figure 4. Composite Shear Stud Connection Details 4. Construction Process The welding of the cap beam elements, stub piles to the cap beam, and shear stud connectors was conducted by the fabricator to eliminate the necessity of any field welding. Hence, only grout placement was necessary to complete the connection at the laboratory. To ensure adequate workability as well strength capacity, an extended working time non-shrining grout product, BASF Masterflow 928, was chosen. Emphasis was also placed in the selection process to ensure that a widely available material was selected. 5

To assist in minimizing the possibility of air voids within the annular grout pocket, the decision was made to pump grout vertically from the bottom of the connection to the top where 1 diameter

6 To assist in minimizing the possibility of air voids within the annular grout pocket, the decision was made to pump grout vertically from the bottom of the connection to the top where 1 diameter holes had been left in the cap beam flange to allow air to escape the connection. A hand operated pumping system was utilized along with shut off valves that were cast into place and later removed to facilitated pumping of the grout as shown in Figure 5 and Figure 6. In addition to the pocketed connections, (12) 4 x8 cylinders were also cast for testing to ensure a minimum compressive strength of 5-6 ksi was obtained by the grout prior to testing of the specimen. Figure 5. Grout Pumping System 6

7 Figure 6. Cast In Place Shut-Off Valve 5. Load History and Instrumentation The test specimen was laterally loaded using a 440 kip MTS hydraulic actuator at North Carolina State University s Constructed Facilities Laboratory as shown in Figure 7. The applied load history, termed a three cycle set history, consisted of an initial elastic portion based on the anticipated yield force of the system and a second section based on the experimentally determined yield displacement of the system. More specifically, single reverse cyclic load controlled cycles of ¼ first yield force increments were applied until a full first yield force cycle was conducted where the first yield force is determined in accordance with Eq. 2. In Eq. 2, S represents the section modulus of the HSS 16x0.500 pile, f y represents the anticipated yield stress of the pile material, and X represents shear span from the pinned supports to the critical section. In the case of test 2, the average material yield strength was found to be 56.5 ksi for the A500 Gr. B pile material based on in-house material testing. This value, combined with the shear span of 8.65 from the pinned supports to the base of the composite connection, resulted in a first yield force of 93.3 kips. The second section of the load history was defined by displacement controlled incremental ductility levels where ductility 1, or effective yield, is defined by Eq. 3 and subsequent displacement ductility levels are defined by Eq. 4. In Eq. 3, Δ y represents the experimentally determined first yield displacement while M p and M y represent the full plastic moment capacity and the first yield 7

8 moment capacity of the pile section respectively. In the case of test 2, the experimentally determined first yield displacement of 2.15 resulted in a ductility 1 displacement of The load and displacement histories resulting from this testing method are provided in Figure 8 and Figure 9 respectively. South Positive Load North Negative Load 440 Kip Actuator Strong Wall Pinned Supports Strong Floor Figure 7. Laboratory Set-Up F y = 2(S)(f y )/(X) Eq. 2 Δ y = (M p /M y )Δ y = µ 1 Eq. 3 µ i = i(µ 1 ) Eq. 4 8

9 Displacement (in) Displacement (cm) Force (kip) Force (kn) F y /4 F y /2 F y μ 1.5 μ 2 μ 1 μ 3 μ 4 μ Figure 8. Test 2 Load History μ 3 μ1.5 μ 2 μ 1 μ 4 μ Figure 9. Test 2 Displacement History 9

10 The instrumentation utilized in test 2 consisted of traditional laboratory instrumentation as well as the Optotrak motion sensing system that tracks the location of LED markers, which was also utilized in phase 1. The traditional equipment consisted of inclinometers located 8 above the center line of the pinned bases to monitor drift magnitudes, linear string potentiometers attached to the bases to monitor any unanticipated base sliding, and strain gauges located on the extreme fibers of each pile cross section as shown in Figure 10. As was mentioned, in addition to the traditional instrumentation, a 2 spaced grid of Optotrak LED markers was placed on the east face of each pile in both the composite connection and critical pile regions as shown in Figure 11. Post processing of the local deformation data recorded by the Optotrak system allows for additional strain calculations over a larger area and at higher levels of strain than is allowed by the traditional electric resistance strain gauges. Figure 10. Strain Gauge Layout Figure 11. Optotrak Marker Grid 6. Experimental Summary The results of experimental testing showed the specimen to perform in an adequate manner with no appreciable signs of damage that could be related to strength loss or any upper bound limit state throughout the second ductility level as is shown in Figure 12 and Figure 13. However, minor signs of inconsequential damage were experienced as early as the first yield cycles where cracking was observed at the base of the grout pocket as shown in Figure 14. Throughout the ductility 1 and 2 cycles, more small cracks developed in addition to the formation of small gaps between the grout block and the pile and/or stub pile walls on the tensile faces of each column as shown in Figure 15. However, this observed damage was not associated any significant recognizable limit state or strength loss. 10

11 Force (kip) Force (kn) Force (kip) Force (kn) Displacement (cm) μ μ μ μ 3 μ μ Displacement (in) Figure 12. Test 2 Force-Displacement Hysteresis Displacement (cm) μ 1.5 μ μ 2 μ 3 1 μ Displacement (in) Figure 13. Test 2 Force-Displacement Envelopes μ 6 Cycle 1 Cycle 2 Cycle

12 Figure 14. North Pile North Face, -Fy Cycle, kips, Small Cracks Figure 15. North Pile South Face, Ductility 2 Cycle 1, 134 kips, 5.62 in 12

The first signs of the development a critical limit state occurred in third positive cycle of the ductility 2 loading stage where a small magnitude of pile wall local buckling began to develop on the

13 The first signs of the development a critical limit state occurred in third positive cycle of the ductility 2 loading stage where a small magnitude of pile wall local buckling began to develop on the north side (compression face) of the south HSS16x0.500 pile along with a small amount of slip between the grout block and the stub pile wall. Upon reversal to the third negative cycle of ductility 2, a similar magnitude of local buckling developed on the south side (compression face) of the north pile as well as grout block slipping as shown in Figure 16. In neither case was the buckling significant enough to produce any noticeable strength loss in either the full force-displacement response or the force-displacement response envelopes. Figure 16. North Pile South Face, Ductility 2 Cycle -3, -132 kips, in Progressing into the first positive and negative cycle of ductility 3, 8.44 of displacement as shown in Figure 17, led to an increase in the magnitude of local buckling on both piles but again no recognizable strength loss in the system. However, the second and third cycles of the ductility 3 level produced propagation of the pile wall local buckling which consequently led to strength reductions of approximately 10% and 15% respectively. Additionally during these cycles, grout located below the first row of shear studs began to spall as shown in Figure

As the test progressed into the ductility 4 cycles, 11.

14 As the test progressed into the ductility 4 cycles, of displacement as shown in Figure 19, intense local buckling developed inducing strength reduction magnitudes of approximately 25% to 40% throughout the ductility level. As the magnitude of the local buckling increased, small cracks began to develop in these regions of the pile wall as shown in Figure 20. As is also shown, grout began to spall to the depth of the first row of shear studs at this ductility level. Although at the completion of the ductility 4 cycles the damage, as well as the associated strength loss, was signification the bent remained in-tact and did not exhibited any complete wall cracking. The test was therefore continued to a ductility level of 6. Figure 17. Ductility 3 Cycle 1, 134 kips, 8.44 in 14

15 Figure 18. Ductility 3 Cycle 2, 126 kips, 8.44in Figure 19. Ductility 4 Cycle 1, 103 kips, in 15

Figure 20. Ductility 4 Cycle 3, North Pile South Face, 85 kips, 11.25 in As the test progressed to the first positive cycle of the ductility 6 level, associated with 16.

16 Figure 20. Ductility 4 Cycle 3, North Pile South Face, 85 kips, in As the test progressed to the first positive cycle of the ductility 6 level, associated with of displacement as shown in Figure 21, pile wall cracking developed in the south face of the north column and propagated until the ductility 6 displacement was reached. Similarly, as the specimen was displaced towards the first negative cycle, pile wall cracking developed on the north face of the south column at the location of prior buckling. This cracking propagated until the ductility 6 displacement was reached as shown in Figure 22. At this point, approximately 60% strength loss had been experienced so the specimen was returned to a neutral position and the test was assumed to be concluded. 16

17 Figure 21. Ductility 6 Cycle 1, 75 kips, in Figure 22. Ductility 6 Cycle -1, -54 kips, in 17

18 Distance Below Cap Beam (in) Following testing, Optotrak data analysis was conducted to evaluate the effectiveness of the connection configuration to reduce strain demands within the connection itself and appropriately relocate damage. As is shown in Figure 23 through Figure 26, the connection did remain in the elastic range while large inelastic demands were forced below the bottom of the connection which terminated at 23 below the cap beam. As was intended in the design of the connection, a critical region developed immediately below the capacity protected connection on both extreme fibers of the south pile in both directions of loading. Strains in the approximate region of (µε) were developed in this critical region prior to the formation of local buckling at the ductility 3 stage of loading. Although not shown in this report for brevity, nearly identical response was experienced on the north pile. It should be noted that in some cases, such as Figure 23, the vertical strain profile appears to enter the positive range on a face that should be experiencing negative strains or vice versa. This condition is generated in a region where significant local buckling has occurred and the orientation of the Optotrak marker is no longer correctly indicative of engineering strains. Micro-Strain (µε) ε y ε y Fy 5 µ1 µ µ2 Cycle 1 µ3 15 µ4 µ6 20 Connection Base Figure 23. Test 2 South Column North Face Positive Cycle 1 Vertical Strain Profile 18

19 Distance Below Cap Beam (in) Distance Below Cap Beam (in) Micro-Strain (µε) ε y ε y Fy 5 µ1 µ µ2 µ3 15 µ4 µ6 20 Connection Base Cycle 1 50 Figure 24. Test 2 South Column South Face Positive Cycle 1 Vertical Strain Profile Micro-Strain (µε) ε y ε y Fy 5 µ1 µ µ2 µ3 15 µ4 µ6 20 Connection Base Cycle 1 Figure 25. Test 2 South Column North Face Negative Cycle 1 Vertical Strain Profile 19

20 Distance Below Cap Beam (in) Micro-Strain (µε) ε y ε y Fy 5 µ1 µ µ2 Cycle 1 µ3 15 µ4 µ6 20 Connection Base Figure 26. Test 2 South Column South Face Negative Cycle 1 Vertical Strain Profile 7. General Conclusions The physical results of experimental testing, as well as the results of data analysis, indicate that the composite connection design considered in this test was effective in producing a desirable failure mode in the form of flexural hinging of the HSS16x0.500 pile wall. Consequently, undesirable failure modes such weld cracking or tearing the pile wall prior to local buckling were avoided. Although some damage did develop within the grout block at the base of the connection in the form of cracking and spalling, no adverse consequences were induced on the structural response due to these actions. Although these actions may be considered as low level response limit states such as a serviceability limit state (hence some repair after seismic loading would be necessary) they would not constitute an ultimate limit state. As has been noted, the ultimate limit state in this case would be considered pile wall local buckling likely corresponding to a reliable displacement ductility level of 3 depending on the exact allowable level of strength loss. It should also be noted that, although associated with large levels of damage and loss of strength, the specimen was able to remain in-tact at displacement ductility of 6 associated with of displacement. This is largely due to the development of a stable failure mode while avoiding more brittle failure mode such as welded connection cracking. From the results presented in this report, the design methodology of developing the axial yield capacity of the pile to determine the required number shear stud connectors appears to be adequate and no optimization options are immediately evident. 20

21 8. Project Direction Following the completion of test 2, progress began on the construction of the test 3 specimen which consists of a semi-dog-bone column capital configuration designed to relocate damage by developing flexural hinging in a reduced section region as shown in Figure 27 and Figure 28. To date, the construction, including all weld inspections, has been completed allowing for instrumentation of the specimen to begin. Provided the anticipated desirable failure mode of flexural hinging in the form of local buckling within the reduced region is shown to control the response of the specimen, the test will serve as validation to and improvement upon a similar design evaluated in test 6 of the first phase of this project. The design evaluated in that test utilized a capital with a single bevel that had no specific reduced section to precisely locate damage. Figure 27. Semi-Dog-Bone Capital Connection Detail 21

22 Figure 28. Semi-Dog-Bone Capital Connection Again, assuming the test 3 specimen performs as anticipated, two viable alternatives to develop the desirable flexural hinging failure will have been developed. These would include the composite shear stud connection discussed in this report and the column capital. The tests following test 3 could include the following options: Test 4: evaluation of composite shear stud connection with construction tolerance offsets (see Figure 29) Test 5: verification of optimal composite shear stud connection if poor performance is exhibited by the evaluation of the construction tolerance offsets Test 5 (or 6) 9: composite connection with other mechanical shear connectors (i.e. bolts) truss system evaluation shake table testing of composite shear stud connection and/or capital connection 22

23 Figure 29. Composite Shear Stud Connection with Construction Tolerance Offset 9. References AASHTO LRFD Bridge Design Specifications, American Association of State Highway and Transportation Officials, 4 th Edition, AISC Specifications for Structural Steel Buildings, American Institute of Steel Construction, March,

Seismic Performance of Steel Bridge Piers Containing Composite Connections

Seismic Performance of Steel Bridge Piers Containing Composite Connections S.J. Fulmer, M.J. Kowalsky & J.M. Nau Department of Civil, Construction and Envinronmental Engineering, North Carolina State University,