Seismic performance of a bridge subjected to far-field ground motions by a Mw 9.0 earthquake and near-field ground motions by a Mw 6.

Transcription

1 i Seismic performance of a bridge subjected to far-field ground motions by a Mw 9. earthquake and near-field ground motions by a Mw 6.9 earthquake REINA GOTO Master of Science Thesis Stockholm, Sweden 212

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3 Seismic performance of a bridge subjected to far-field ground motions by a Mw 9. earthquake and near-field ground motions by a Mw 6.9 earthquake Reina Goto June 212 TRITA-BKN. Master Thesis 358 ISSN ISRN KTH/BKN/EX-358-SE

4 Reina Goto, 212 Royal Institute of Technology (KTH) Department of Civil and Architectural Engineering Division of Structural Engineering and Bridges Stockholm, Sweden, 212

5 Preface This Master s thesis was initiated with the help of Kawashima Research Group at the Department of Civil Engineering at the Tokyo Institute of Technology, Tokyo Tech, and was carried out at the Department of Civil and Architectural Engineering at the Royal Institute of Technology, KTH. First and foremost, I would like to give my sincere gratitude to Professor Kazuhiko Kawashima, Tokyo Tech, and Assistant Professor Hiroshi Matsuzaki, Tohoku University. I would like to thank Professor Kawashima for his great support and advices. I would like to thank Assistant Professor Matsuzaki for helping me with all my questions and to understand the seismic design methods and the dynamic response analysis better. I would like to thank my supervisor Post Doctor Nora Ann Nolan, KTH, for her great support and giving me valuable advices. I would also like to thank my examiner Professor Raid Karoumi, KTH, for showing great interest and being very helpful during the course of the thesis. Special thanks to the members of the Kawashima Research Group, Tokyo Tech, for all their help and for sending me papers and files needed for the research. Stockholm, June 212 Reina Goto i

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7 Abstract In the last two decades, two major earthquakes have occurred in Japan: the 1995 Kobe earthquake and the 211 Great East Japan earthquake. In the 211 Great East Japan earthquake, many bridge structures were destroyed by the tsunamis, but it is interesting to study the ground motion induced damage and also how this earthquake differed from the one in In this thesis, the seismic response of a bridge designed according to the current Japanese Design Specifications was evaluated when it is subjected to near-field ground motions recorded during the 1995 Kobe earthquake and far-field ground motions recorded during the 211 Great East Japan earthquake. For this purpose, a series of nonlinear dynamic response analysis was conducted and the seismic performance of the bridge was verified in terms of its displacement and ductility demand. It was found from the dynamic response analysis that the seismic response of the target bridge when subjected to the ground motions from the 211 Great East Japan earthquake was smaller than during the 1995 Kobe earthquake. Although the ground motions from the 211 Great East Japan earthquake were very strong, they were not as strong as the ground motions from the 1995 Kobe earthquake. The results obtained in this thesis clarify the validity of the Type I and Type II design ground motions. The target bridge used in this thesis was designed according to the post-199 design specifications and showed limited nonlinear response when subjected to the different ground motions which shows how efficient the enhancement of the seismic performance of bridges has been since the 199 s. Keywords: seismic performance, dynamic response analysis, far-field ground motions, near-field ground motions iii

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9 Notations Main notations A h A w b c c c D c e c pt c R c s c Z d D d R d Ra Sectional area of each lateral confining reinforcement Sectional area of reinforcing bars Width of the column section Cyclic loading effect factor Damping modification factor Effective height factor Modification factor depending on the longitudinal tensile reinforcement ratio Factor depending on the bilinear factor Response modification factor Zone modification factor Effective length of lateral confining reinforcement Effective height of the column section Residual displacement developed at a column Allowable residual displacement at a column d Ra,LG Allowable residual displacement in the longitudinal direction d Ra,TR Allowable residual displacement in the transverse direction d u d y E c E des f c f cc f ck f sy Ultimate displacement of column Yield displacement of column Young s modulus of concrete Descending gradient Strength of concrete Strength of confined concrete Design strength of concrete Yield strength of reinforcement bars gal Measure of acceleration (1 gal = 1 cm/s 2 ) h Height of the column/effective height of the column section v

10 k hc k hc L p M y M u P P a P s P s P u r s S S S I S II S c S i S s T T i W W p W U c cc s sy a Standard modification coefficient Design horizontal seismic coefficient Plastic hinge length of column Initial yield moment Ultimate moment Lateral strength Lateral capacity of a column Shear strength of a column Shear strength under static loading of a column Ultimate lateral strength of a column Bilinear factor Spacings of lateral confining reinforcement Response acceleration spectrum for the Level 1 earthquake ground motion Standard acceleration spectra for Level 1 earthquake ground motion Response acceleration spectra for Type I ground motion Response acceleration spectra for Type II ground motion Shear capacity resisted by concrete Standard acceleration response spectra Shear capacity resisted by transverse reinforcement Fundamental Period Natural periods Equivalent weight Weight of the column Weight of part of the superstructure supported by the column concerned Shape factor/safety factor Shape factor Strain of concrete Strain of concrete under the maximum compressive stress Volumetric ratio of lateral confining reinforcements Yield point of the reinforcements Damping ratio Design displacement ductility factor of a column vi

11 R c u y Response displacement ductility factor of a column Average shear stress that can be borne by concrete Ultimate curvature Yield curvature Abbreviations AIS Arc Information Systems EW East-West horizontal component of ground motion JRA Japan Road Association LG Longitudinal NIED National Research Institute for Earth Science and Disaster Prevention NS North-South horizontal component of ground motion PGA Peak ground acceleration RC Reinforced concrete SPL Seismic Performance Level TR Transverse UD Up-Down vertical component of ground motion WSJ The Wall Street Journal vii

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13 Contents Preface... i Abstract... iii Notations... v 1 Introduction Aim and scope of thesis Organization of thesis Background and previous studies Seismic history of Japan Kobe earthquake Shear failure of RC columns Collapse of steel columns Damage to unseating prevention devices Damage to steel bearings Great East Japan earthquake Bridges designed before Bridges that had been retrofitted or designed after History of seismic design of bridges in Japan Current seismic design Basic principles Analytical methods to verify the seismic performance Design of RC columns Previous studies Methodology Ground motions Response acceleration spectra Target bridge General ix

14 3.3.2 Reinforced concrete columns Design details of the RC columns Design process of the columns Bearings Foundations Finite element analysis program TDAP III Analytical idealizations Mass idealizations Damping idealizations Structural elements Fiber elements Linear springs Material properties Analysis using TDAP III Self-weight analysis Eigen value analysis Dynamic response analysis Sensitivity analysis Convergence study Time step Number of fiber elements Results Mode shapes and natural period of the bridge Comparison between the 1995 and 211 earthquake Relative response displacement at the top of column Relative response displacement at the deck Moment vs. curvature hysteresis Stress vs. strain hysteresis Verification of seismic performance Ductility capacity and demand Residual displacement Conclusions and suggestions for further research Conclusions Suggestions for further research x

15 References Appendix A Ground motions UD Appendix B Calculations RC column Appendix C Mode shapes xi

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17 Chapter 1 Introduction Japan is situated on a region where several tectonic plates meet, which is why Japan is extremely prone to earthquakes. There have been many earthquakes in the past and many lessons to be learnt alongside it. Japan has made huge investments to improve buildings and infrastructures to mitigate seismic damage. The Japanese seismic design codes have been revised several times and revisions are sure to appear in the future. On March 11 th 211, Japan was hit by a huge earthquake called 211 Great East Japan earthquake. It was the biggest earthquake ever recorded in Japan and it was apparent that the country was not prepared for the kind of damages that followed the earthquake. Not only did this earthquake cause immense damage and casualties, but it also caused the biggest nuclear disaster since Chernobyl in 1986 to further grieve the people of Japan. Many bridge structures were destroyed by the tsunamis, but it is interesting to see how the ground motions of the earthquake damaged these bridge structures. In the context of seismic design of bridges, perhaps there are lessons to be learnt from this earthquake. To mitigate seismic damage of bridges, it is important to find out how this earthquake differed from other earthquakes in the past and whether or not the Japanese Seismic Design Specifications for bridges need to be revised. 1.1 Aim and scope of thesis The aim of this thesis is to evaluate the seismic response of a bridge designed by the current Japanese seismic design codes when it is subjected to ground motions recorded during the two major earthquakes that have occurred in Japan in the last two decades: the 211 Great East Japan earthquake and the 1995 Kobe earthquake. For this purpose, a series of nonlinear dynamic response analysis of a bridge is conducted. The seismic performance of the bridge is then verified in terms of its displacement and ductility demand. First, the seismic history of Japan will be studied to understand the damages that have occurred in the past. The 1995 Kobe earthquake and the 211 Great East Japan 1

18 earthquake will be studied in more detail, since the ground motion records from these earthquakes will be used in this thesis. The current Design Specifications and how it has changed over the years since its first publication will also be studied. A literature review to find any information that is relevant to this thesis will be conducted and presented. Ground motion records from the 1995 Kobe earthquake and the 211 Great East Japan earthquake will be evaluated to see differences in the ground motion characteristics. Also response acceleration spectra for the ground motion records will be analyzed to see and compare the intensity and predominant period of each ground motion. A bridge based on the Japanese Seismic Design Specifications is used to conduct a dynamic response analysis using a Japanese finite element analysis program called TDAP III. The seismic response and seismic performance of the bridge when subjected to different ground motions will then be evaluated. General steps in this thesis are: 1. Evaluate how the two earthquakes differ in character. The ground motion characteristics will be compared as well as its response acceleration spectra. 2. Conduct dynamic response analysis using TDAP III. 3. Compare the seismic response of the bridge and evaluate its seismic performance based on the results from the dynamic response analysis. 4. Discuss whether or not the current Japanese Seismic Design Specifications for bridges are sufficient for an earthquake with a different character than the 1995 Kobe earthquake such as the 211 Great East Japan earthquake. Several assumptions and simplifications were made in this study. The target bridge was taken from an example book issued by the Japan Road Association and it was designed based on nonlinear static analysis. In reality nonlinear dynamic response analysis should be conducted when designing a bridge, but in this case the bridge was designed based on only nonlinear static analysis for simplicity. In the dynamic response analysis, the difference in the arrival time of the earthquake ground motions were not considered since the length of the target bridge is only.2 km. The difference in the arrival time should be considered for longer bridges. Also, torsion and shear deformation were not considered in this analysis. The damping ratios of the elastomeric bearings, soil springs, and structural components of the bridge were assumed using values from the Japanese Design Specifications (JRA, 22). These damping ratios were assumed since the aim of this study is to compare the seismic response of the target bridge when subjected to 2

19 different ground motions and to find a precise damping is not of interest. Damping of the bridge structure was idealized using Rayleigh damping (see Section for the calculations of the Rayleigh coefficients and the damping curve). 1.2 Organization of thesis Chapter 2 gives some background information necessary for this thesis. A short summary of the seismic history of Japan is presented and the 211 Great East Japan earthquake and 1995 Kobe earthquake are presented in detail. The history of seismic design in Japan is summarized linking the revisions of the Seismic Design Specifications to damages that were observed after some of the major earthquakes. Also the current Japanese Seismic Design Specifications are presented. Previous studies of relevance are discussed. The methods of analysis, the target bridge, and the bridge model are presented in detail in Chapter 3. The results of the analysis are presented in Chapter 4 and are analyzed. In Chapter 5, the conclusions that were deduced from this study are presented and any suggestions for future research are discussed. 3

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21 Chapter 2 Background and previous studies 2.1 Seismic history of Japan Japan has a long history of earthquakes and some of the more significant earthquakes, in terms of seismic design, will be presented here. In the early 19 s when seismic effects were either not or poorly considered in design, the 1923 Kanto earthquake with a moment magnitude of 7.9 occurred in the Tokyo-Yokohama area (Kawashima, 2). This earthquake caused large scale damage to buildings and infrastructure, where bridges collapsed due to tilting, overturning, and settlement of the foundations. Due to this earthquake, the importance of considering seismic effects in design was recognized for the first time (Kawashima, 211). In 1964, an earthquake with a moment magnitude of 7.5 occurred in Niigata which came to be called the 1964 Niigata earthquake. Many bridges were damaged or collapsed due to soil liquefaction and it was at this time that the actual term liquefaction was first coined (Kawashima, 211). It became evident after this earthquake that soil liquefaction needed to be considered in seismic design. However, at that time, further research to understand the mechanism of liquefaction was needed before implementing any countermeasures. Bridges were also damaged by large relative displacements of the decks, which inspired the development and implementation of unseating prevention devices. After the 1964 Niigata earthquake and up to the early 199 s, several big earthquakes occurred. However, the damages in these earthquakes were quite limited due to changes in seismic design practices. It was not until 1995, that a big earthquake that would greatly change the seismic design practices in Japan occurred. This was the 1995 Kobe earthquake and it had a great impact on the seismic design of bridges. Even to this day, ground motions from this earthquake are used for dynamic response analysis of bridges. Since the ground motions from the 1995 Kobe earthquake are used in this thesis, it is presented more in detail in Section Similarly, the recent 211 Great East Japan earthquake is presented in detail in Section 2.1.2, since ground motions 5

22 from this earthquake are also used in this thesis and the damages that occurred need to be thoroughly described Kobe earthquake In 1995, 17 th of January, the 1995 Kobe earthquake occurred at Kobe and Awaji Island, in southern Japan. This earthquake had a moment magnitude of 6.9 and near-field ground motions were recorded. Thousands of deaths and extensive damage to buildings and infrastructures were reported. Many bridges suffered damage, where 9 highway bridges collapsed or nearly collapsed and 16 bridges were severely damaged. The four major types of damages that were observed are summarized below, based on the lecture notes from Seismic design of urban infrastructure (Kawashima, 211) Shear failure of RC columns Figure 2.1 shows the collapse of the 18-span Fukae Viaduct of the Hanshin Expressway in Kobe. This viaduct was designed based on the 1964 Design Specifications that will be presented later in Section 2.2. During the earthquake, the RC columns which were 9.9 m to 12.4 m tall with a diameter of 3.1 m to 3.3 m were damaged by severe flexural and diagonal cracks that developed 2.5 m above the footing. This was where one third of the longitudinal reinforcement bars terminated. Since the amount of tie bars were not enough, premature shear failure shown in Figure 2.2 also occurred in the columns. These damages occurred due to the deficiencies in design. For instance, the allowable shear stress was overestimated and the development length of the longitudinal bars was insufficient. This kind of failure occurred in several other bridges as well such as in the Takashio Viaduct which was built according to the 1971 Design Specifications. Figure 2.1: The collapse of the Fukae Viaduct. (Kawashima, 211) 6

23 Figure 2.2: Premature shear failure of column of the Fukae Viaduct. (Kawashima, 211) Collapse of steel columns Steel columns collapsed in numerous bridges and an example of this is Tateishi Viaduct at the Hanshin Expressway. A picture of a collapsed column from this viaduct is shown in Figure 2.3. This viaduct was also built based on the seismic coefficient method from the 1964 Design Specifications. The steel columns were built between two RC columns at the sides and lateral beams were constructed to support two side decks. To reduce damage of the steel columns in the event of an automobile accident, the inside of the columns were filled with weak concrete from the bottom up to a height of 2.3 m. During the earthquake, local buckling of web and flange plates and rupture of the welded corners at the bottom of the columns occurred. This caused the bearing capacity of the columns to decrease in the lateral and vertical directions. The columns became vulnerable to the dead weight of the decks and started to settle. When this happened, the decks in the center started to buckle and in the end the steel columns collapsed. Figure 2.3: Collapse of a steel column of the Tateishi Viaduct. (Kawashima, 211) 7

24 Damage to unseating prevention devices Damage to various types of unseating prevention devices was observed. This happened since the design force of the devices was too small. The design force was calculated by multiplying the static reaction force by a seismic coefficient of.3 to.4. In the Nishinomiya Bridge of the Hanshin Expressway, one of the approach spans collapsed (Figure 2.4.a). The main bridge and the approach spans were connected by plate-type restrainers (Figure 2.4.b). During the earthquake, the fixed bearings of the main bridge failed and caused the bridge to displace, pulling the approach span. Eventually the approach spans dislodged from its supports and collapsed since the unseating prevention devices could not support it without the help of the supports. a) Collapse of an approach span b) Failure of a plate-type restrainer (Nishinomiya Bridge) Figure 2.4: Damage to unseating prevention devices (Kawashima, 211) Damage to steel bearings Extensive damage to steel bearings was also observed in this earthquake (Figure 2.5). Prior to the 1995 Kobe earthquake, steel bearings were thought to restrict extensive damage to the bridge substructures. However, after observing the damage caused by the failure of steel bearings, it became apparent that steel bearing were one of the main causes of the extensive damage that occurred (Kawashima, 211). This is because steel bearings are weak for shock and have insufficient strength and length of movement. Apart from the three above mentioned damages, damage to bridge foundations were also observed. However these damages were minor compared to the rest of the structural components. Damage caused by soil liquefaction was also observed in the form of settlements and tilting of foundations and bridge substructures. Foundations were also damaged by large lateral spreading which was caused by soil liquefaction. The Japanese Design Specifications were revised in 1996 due to the poor seismic performance of bridges in this earthquake. The revisions that were made in the Design Specifications will be presented in Section

25 a) Failure of steel pin bearing b) Failure of steel bearing Figure 2.5: Failure of steel bearings (Kawashima, 211) Great East Japan earthquake On March 11, 211 a devastating earthquake of moment magnitude 9. occurred off the Pacific coast, northeast of Japan. This was the biggest earthquake ever recorded in Japan and was named the 211 Great East Japan earthquake. This earthquake lasted for more than 3s and strong ground motion accelerations were recorded in several areas. The coastal regions of northeast Japan were hit by tsunamis after the earthquake which caused severe damage to buildings and infrastructures, human injuries, and casualties. The earthquake was felt all the way down to the Kanto region and extensive soil liquefaction occurred in the Tokyo Bay area as well as in Chiba Prefecture where damage such as settlements of buildings and uplift of sewage manholes were observed (Ishihara, 212). The tsunamis swept away and damaged several bridges along the coast, but damage to bridges which was induced by ground motions was less extensive. However, according to a study by Kawashima et al. (211) and Kawashima (212), bridges that were designed based on the design codes prior to the 199 and 1995 Design Specifications and were not retrofitted were damaged due to the ground motions. Bridges that had been retrofitted or built according to the post 199 Design Specifications showed only minor damage or no damage at all. This showed that seismic retrofitting and the improvements that had been made in the Design Specifications were efficient. The bridge damage that was observed after the 211 Great East Japan earthquake is presented within two categories: bridges designed pre-199 and post

26 Bridges designed before 199 The same type of damage to RC columns as in the 1995 Kobe earthquake occurred, which was mentioned in Section In the Esaki Ohashi Bridge, damage to the RC columns was observed which can be seen in Figure 2.6.a. This type of damage occurred in bridges that were designed before the 199 s and had an overestimated shear capacity and not enough development length of the longitudinal bars. Kunita Ohashi Bridge was also designed prior to the 199 s and had not been retrofitted at the time of the earthquake. This bridge was closed for service after the earthquake, since its steel bearings were damaged (Figure 2.6.b) and shear cracks had developed in the RC columns. The information on the damage on the Esaki Ohashi Bridge and the Kunita Ohashi Bridge were obtained from a study by Hoshikuma et al. (212). a) RC columns b) Steel bearings (Esaki Ohashi Bridge) (Kunita Ohashi Bridge) Figure 2.6: Damage to bridges designed prior to 199-design codes (Hoshikuma et al., 212) Bridges that had been retrofitted or designed after 199 Bridges that had been retrofitted after the 1995 Kobe earthquake, by for example steel jacketing of RC columns and replacing steel bearings with elastomeric bearings, showed in most cases no signs of damage. Bridges that were designed according to the post-199 Design Specifications were not damaged or showed only minor damages. However, some bridges suffered severe damage to its elastomeric bearings and dampers. One of these bridges was the Tobu Viaduct in Sendai, where elastomeric bearings ruptured (Kawashima, 212). Figure 2.7.a show how the rupture of the bearings caused the bridge deck to offset in the transverse direction by.5 m and Figure 2.7.b show that the rubber layers detached from the steel plates and ruptured. Some possible reasons to why this damage occurred could be because of a design miss or that the interaction of adjacent decks was not properly considered (Takahashi, 212 and Kawashima, 212). Damage was also observed in the attachments and anchors of dampers (Figure 2.8). 1

27 NEXCO East a) Offset due to rupture of b) Rupture of elastomeric bearing elastomeric bearings Figure 2.7: Damage of elastomeric bearings in the Tobu Viaduct (Kawashima, 212). a) Damage of anchors b) Damage of attachment Figure 2.8: Damage of attachments and anchors of dampers (Takahashi, 212). 2.2 History of seismic design of bridges in Japan The revisions and history of the Japanese Design Specifications for seismic design of bridges will be presented in this subsection, based on lecture notes from Seismic Design of Urban Infrastructures (Kawashima, 211) and papers by Professor Kawashima (Kawashima, 2 and Kawashima, 26). In 1926, three years after the 1923 Great Kanto earthquake, the first Japanese seismic provisions for highway bridges were published. In these specifications, the seismic coefficient method using a seismic coefficient of.1 to.3 was included and only the requirement of seismic lateral force of 2% gravity force was presented. Design specifications of steel highway bridges were included in 1939 and were revised twice afterwards in 1956 and At this time, earthquake engineering was still something new and under progress, so the seismic design requirements in these specifications were far from what they are now. It was not until the 1964 Niigata earthquake that engineers realized the need for major improvements of the seismic provisions. After 11

28 observing the damages caused by the 1964 Niigata earthquake, a completely renewed seismic design provisions, Guide Specifications for Seismic Design of Highway Bridges, were issued in Some of the improvements and changes that were made are presented below: The lateral force should be calculated by considering the zone, importance of the bridge, and ground condition in the seismic coefficient method. Also the structural response should be considered in the modified seismic coefficient method. Since many bridges were damaged by soil liquefaction in the 1964 Niigata earthquake, the evaluation of soil liquefaction was included. However, the mechanism of soil liquefaction was unknown at that time so design procedure for liquefaction could not be included in The need for unseating prevention devices were recognized in this earthquake so several types of unseating prevention devices such as steel plate connectors and cable restrainers were developed. Many independent methods for the design of substructures had been developed and these methods were unified as Guide Specifications of Substructures between 1964 and This resulted in the development of new types of foundations which helped reduce the damage of the bridge foundations. In 198, the above Guide Specifications for seismic design and substructures were revised. These specifications were written as Part V Seismic Design and Part IV Substructures in the Design Specifications of Highway Bridges. Parts I to III were the General Aspects, Steel Bridges, and Concrete Bridges respectively. A method for the design of foundations in liquefying soils and an updated version of the evaluation method for predicting soil liquefaction were added in Part V. In Part IV, the allowable shear stress for concrete was reduced since this was overestimated in the past. The anchoring length of the reinforcement bars from the footings was increased to 2 times the diameter of the bars and the length equivalent to the effective width of the column. The Design Specifications were revised again in 199. In this revision, the following changes were made: The seismic coefficient method and the modified seismic coefficient method were unified. For the first time, to enhance the ductility of bridge columns, the check of the strength and ductility of the reinforced concrete columns was included. The nonlinear behavior of a bridge was to be checked after the structural members yielded. The Type I ground motion of the standard lateral force coefficient in 12

29 Figure 2.9 was introduced for the ductility check. This ground motion represents the ground motions that are assumed to have occurred in the 1923 Kanto earthquake. Type II ground motion was included in the later revisions. The static frame method was introduced so that the lateral force distribution of a multi-span continuous bridge could be evaluated. Through this method, the three dimensional behavior of a bridge could be considered in the equivalent static analysis. Figure 2.9: The standard lateral force coefficient (Kawashima, 2). As previously mentioned, even though strong earthquakes occurred several times in the 198 s and the beginning of 199 s, the damages were quite limited due to the improvements that had been made in seismic design. Therefore, the damages that resulted from the 1995 Kobe earthquake were somewhat shocking. 4 days after this earthquake, the Guide Specifications for reconstruction and repair of highway bridges which suffered damage in the 1995 Kobe earthquake was issued to guide the reconstructions of the bridges that were damaged in this earthquake. This Guide Specifications came to be used in new constructions of bridges as well, until a revised version of the Design Specifications came out in In this Guide Specifications, a requirement for the design of a plastic hinge at the bottom of columns and the effect of lateral confinement was included. Also, the Type II ground motion in Figure 2.9 was included which represents the ground motions recorded in the 1995 Kobe earthquake. In 1996, the Design Specifications from 199 were fully revised and included the above mentioned 1995 Guide Specifications. Some of the major changes that were made are the following: The previous check of the ductility of the reinforced concrete columns was improved to the ductility design method. Although the seismic coefficient method was still in use, revisions in the Design Specifications were made so that all the structural components that are vulnerable to seismic effects are to be checked with the ductility design method. 13

30 The type of ground motion (Type I and Type II) is to be considered in determining the design ductility factor and shear strength of a bridge column, and also in determining the soil strength for liquefaction. Specifications for the dynamic analysis were revised, where revisions were made in the analytical models and methods, and safety checks. Also the input earthquake ground motions to be used in dynamic analysis were specified. Requirements for the residual displacement of a column after an earthquake were included and this had to be checked for bridges in the important bridge category. An unseating prevention system was introduced and design loads and methods were specified. The function of the unseating prevention devices was also clarified. Elastomeric bearings were recommended to be used as opposed to steel bearings which have several deficiencies. The seismic design treatment of soil liquefaction was reviewed and is to be used as a seismic design method in places where liquefaction is likely to occur. The seismic design treatment of lateral spreading caused by soil liquefaction was also defined. Since 1996, the Design Specifications have been revised in 22. Revisions were made based on the Performance-based design concept, where requirements of the necessary performance and verification of policies are clearly stated. Some of the changes that were made are summarized in the following points: Seismic performance requirements of highway bridges, principles of seismic performance verifications, and the determination concept of design earthquake ground motion were clearly defined. These specifications were based on concepts from the performance-based design. The methods of verifying seismic performance were rearranged to two design methods: Static analysis and Dynamic analysis. The verification method for the latter analysis was defined in detail and its applicability was improved. A method to verify the seismic performance of abutment foundations on liquefiable grounds was included for the first time. Similarly, a method to verify the seismic performance of steel and concrete superstructures was introduced. 14

31 The current Design Specifications will be presented in detail below in Section 2.3. The Design Specifications have been revised again in March 212, but this has not been published yet. Therefore the revisions that were made in 212 will not be discussed in this study and the Design Specifications from 22 will be used. 2.3 Current seismic design In this section, the Design Specifications from 22 (JRA, 22) will be presented. The Design Specification of Highway Bridges is issued by the Japan Road Association (JRA) and consists of five parts: Part I Common, Part II Steel Bridges, Part III Concrete Bridges, Part IV Substructures, and Part V Seismic Design. Some key parts of the Part V Seismic Design will be presented in this section based on lecture notes from Seismic Design of Urban Infrastructures (Kawashima, 211), papers by Professor Kawashima (Kawashima, 24 and Kawashima, 26), and the English translation of the Part V Seismic Design by JRA (JRA, 22) Basic principles In seismic design, a bridge must be designed so that its required seismic performance is satisfied during an earthquake. The seismic performance of a bridge is determined by the importance of the bridge and also the levels of design ground motion that is likely to occur at the site of construction. Furthermore, the topographical-, geological-, soil-, and site conditions must be considered in seismic design. Table 2.1 shows the seismic performance matrix. Bridges are categorized into two types; either Type A or Type B. Type A are bridges with standard importance and Type B are bridges with high importance. The importance of the bridge is classified by using Table 2.2. The type of design ground motions is divided into two levels: Level 1 Earthquake which is ground motions with a high probability occurrence and the Level 2 Earthquake which is ground motions with a low probability occurrence. The design response acceleration spectra for these design ground motions can be seen in Figure 2.1. The Level 1 Earthquake is the ground motions that are developed in moderate earthquakes and the ground motion used in conventional elastic design method. The Level 2 Earthquake includes two types of ground motions: Type I and Type II. Type I represents ground motions developed in interplate-type earthquakes with a large magnitude, which targets the ground motions that most likely occurred in the 1923 Kanto earthquake. Type II represents ground motions developed in inland-nearfieldtype earthquakes and the ground motions from the 1995 Kobe earthquakes are typical targets of this type. Type I ground motion is characterized as having a large amplitude and longer duration, while Type II is characterized as having strong accelerations and shorter duration. 15

32 Depending on the bridge type and design ground motions, the Seismic Performance Level (SPL) needs to be ensured. SPL 1 requires bridge damage to be prevented, which means that the main functions of the bridge must be maintained during an earthquake. SPL 2 requires limited damage in order to recover its function, meaning that the bridge should only suffer limited damage and be able to recover within a short time. In SPL 3, critical damage of the bridge must be prevented. a) Level 1 Earthquake b) Level 2 Earthquake (Type I) c) Level 2 Earthquake (Type II) Figure 2.1: Design acceleration spectra (JRA 22, Kawashima 24). 16

33 Table 2.1: Classification of importance of bridges (JRA, 22). Type A bridges B bridges Definitions Bridges other than Type B bridges Bridges of National expressways, urban expressways, designated city expressways, Honshu-Shikoku highways, and general national highways. Double-deck bridges and overbridges of prefectural highways and municipal roads, and other bridges, highway viaducts, etc., especially important in view of regional disaster prevention plans, traffic strategy, etc. Table 2.2: Seismic performance matrix (JRA, 22). Type of Design Ground Motions Level 1 Earthquake: Ground Motions with High Probability to Occur Standard Bridges (Type-A) Important Bridges (Type-B) SPL 1: Prevent Damage Level 2 Earthquake: Ground Motions with Low Probability to Occur Interplate Earthquakes (Type-I) Inland Earthquakes (Type-II) SPL 3: Prevent Critical Damage SPL 2: Limited Damage for Function Recovery The loads and load combinations that need to be considered in the seismic design of bridges are the primary and the secondary loads. These loads are shown below in Table 2.3. The combination of the loads should be: primary loads + effects of earthquake. The loads and its combinations should be determined to give the most unfavorable condition. Depending on the site of construction, not all loads will be considered. According to JRA, the live load does not need to be considered in seismic design. This is because the live load varies temporally and spatially and during an earthquake, the probability of a full live load occurring is small. 17

34 Table 2.3: Primary and secondary loads to be considered in design (JRA, 22). Primary loads Dead load Prestress force Effect of creep of concrete Effect of drying shrinkage of concrete Earth pressure Hydraulic pressure Buoyancy or uplift Secondary loads Effects of earthquake The effects of earthquake include: Inertia force due to the dead weight of the structure Earth- and hydrodynamic pressure during an earthquake Effects of liquefaction and liquefaction-induced ground flow Ground displacement during an earthquake Analytical methods to verify the seismic performance In the Japanese Design Specifications, to verify the seismic performance of a bridge, the limit state of each structural member should be defined considering the limit states of the bridge. If the response of the structural members due to the design ground motions does not exceed the determined limits, the seismic performance is verified. The limit states of the bridge are the Seismic Performance Levels 1, 2, and 3 which were briefly mentioned in Section These limit states are determined considering the requirements summarized in Table 2.4 from the Design Specifications. Table 2.4: Establishing the Seismic Performance Levels (JRA, 22). Seismic Performance Level SPL 1 SPL 2 SPL 3 Limit States Mechanical properties of the bridges maintained within the elastic ranges Only the structural member in which the generations of plastic hinges are allowed deforms plastically within a range of easy functional recovery Only the structural member in which the generations of plastic hinges are allowed deforms plastically within a range of the ductility limit of the member The design earthquake ground motions, and structural type and limit states of the bridge must be considered when choosing the appropriate analytical method to verify 18

35 the seismic performance. The appropriate analytical method is either a static or dynamic analysis. For a proper evaluation of the seismic performance, the nonlinear behaviors of a member might need to be considered so an appropriate analytical method must be chosen to account for these properties. See Table 2.5 for the required analytical method depending on the complexity of seismic behavior and the SPL s. When determining the seismic performance by a static analysis, the loads that are caused by an earthquake are added statically to the bridge. The dynamic structural characteristics in the elastic range are considered in the seismic coefficient method when verifying for SPL 1. In the seismic coefficient method, loads that have been calculated by using the seismic coefficient are applied to the bridge statically. From this, the resultant deformations and sectional forces are evaluated. In ductility design method, the deformation properties and dynamic strength of the nonlinear zone of a structure are considered. This method is used for the verification of SPL 2 and SPL 3. In both the seismic coefficient method and design ductility method, the dynamic seismic forces are changed to a static force by using the seismic coefficient. When a dynamic method is used for seismic performance verification, the maximum response values of the bridge obtained from the dynamic analysis must be smaller than the allowable values. The response spectrum or time-history response analysis methods are commonly used in dynamic analysis. The most suitable method and model are chosen considering the purpose of the analysis and the earthquake ground motion level. 19

36 Table 2.5: Relation between Complexity of Seismic Behavior and Design Methods Applicable to Seismic Performance Verification (JRA, 22). Dynamic characteristics of bridges Seismic Performance to be verified SPL 1 SPL 2 & SPL 3 Examples of applicable bridges Bridges without complicated seismic behavior Static analysis Static analysis Other than bridges shown in the right columns Bridges with plastic hinges & yielded sections, and bridges not applicable of the Energy Conservation Principle Static analysis Dynamic analysis Bridges with rubber bearing to disperse seismic lateral forces Seismically-isolated bridges Reinforced Concrete rigid-frame bridges Bridges with steel piers likely to generate plastic hinges Bridges of likely importance of higher modes Dynamic analysis Dynamic analysis Bridges with long natural periods Bridges with high piers Bridges not applicable of the Static Analysis Methods Dynamic analysis Dynamic analysis Long-span bridges such as cablestayed bridges and suspension bridges Deck-type & half throughtype arch bridges curved bridges Design of RC columns RC columns are designed so that it satisfies the following requirement in Equation 2.1. P a k hc W (2.1) W W U c W p p (2.2) where, P a is the lateral capacity of a column, k hc is the design horizontal seismic coefficient, W is the equivalent weight, W U is the weight of part of the superstructure supported by the column concerned, W p is the weight of the column, and c p is the equivalent weight coefficient (.5 for bending failure or shear failure after flexural yielding and 1. for shear failure). 2

37 The design horizontal seismic coefficient is calculated using Equation 2.3: k hc csczk. hc 4c z (2.3) where, c s is the response modification factor, c z is the zone modification factor (=.7,.85, or 1. depending on the zone), and k hc is the standard modification coefficient. The response modification factor, needed to calculate the above mentioned requirement, may be calculated using Equation 2.4, which assumes the equal energy principle. The equal energy principle is more conservative than the equal displacement principle. c S 1 (2.4) 2 1 a where, a is the design displacement ductility factor of a column. In order for the RC column to perform according to its expected seismic performance, the response displacement ductility factor,, should be smaller than the design displacement ductility factor, a r. However it may not be greatly smaller, since this would result in an overestimation of the response modification factor. A plastic hinge, which can show ductile behavior when it is subjected to repeated alternate deformations, can be defined at the bottom of each RC column. The plastic hinge region dissipates energy through plastic deformation without collapsing the remaining structural members and by designing these plastic hinges in a proper way can allow the damage that occurs after an earthquake to be localized and repaired more easily (Long and Bergad, 24). The plastic hinge length is determined in the Japanese Design Specifications using Equation 2.5, however it must be in the interval.1d LP. 5D. In analytical purposes, the plastic hinge is a virtual concept which allows the displacement due to plastic deformation at the defined plastic hinge region to be evaluated more easily (Kawashima, 211). L P.2h. 1D (2.5) where, h is the height of the column and D is the effective height of the column section. For every column which has a defined plastic hinge, a fiber element analysis is performed at the plastic hinge regions assuming a stress vs. strain relationship for concrete and reinforcing bars. An elastic-perfect plastic model is used to idealize the stress vs. strain relationship of reinforcing bars. The stress vs. strain relationship of 21

38 confined concrete is based on Hoshikuma et al. (1997) which is presented below in Equation 2.6 to Equation f c n 1 c E c c 1 n cc fcc Edes c cc Ec cc E f c cc cc n1 c cc (2.6) cc c cu (2.7) f cc f ck 3.8 f s sy (2.8) s f cc f E des 2 fck 11.2 f s sy ck sy (2.9) (2.1) s 4A h sd.18 (2.11) where, f c is the strength of concrete, f cc is the strength of confined concrete, f ck is the design strength of concrete, f sy is the yield strength of reinforcement bars, c is the strain of concrete, cc is the strain of concrete under the maximum compressive stress, E c is the Young s modulus of concrete, E des is the descending gradient, and are shape factors, and s is the volumetric ratio of lateral confining reinforcements, A h is the sectional area of each lateral confining reinforcement, and s and d are the spacings and effective length of lateral confining reinforcement. The shape factors are obtained by the following Table 2.6. Table 2.6: Shape factors for circular and rectangular columns. Circular Rectangular According to the Design Specifications, the volumetric ratio of lateral confining reinforcements should be smaller than 1.8%. This recommendation was proposed since there must be a limitation to how much the ductility capacity of a column should be enhanced by just increasing the amount of lateral reinforcement. If the restraining force of concrete is too high, the plastic hinge region will generally become smaller when the column is subjected to repeated plastic deformations. This can cause the longitudinal reinforcements to fracture leading the column to reach the ultimate state. 22

39 The ultimate displacement, d u, is defined as the displacement at the gravity center of a superstructure when the compression strain of the concrete at the out-most reinforcements reaches the ultimate strain, cu, in Equation The ultimate strain is dependent on the type of ground motion. cu cc cc.2 f E des cc Type I ground motion Type II ground motion (2.12) The ultimate displacement of a column, d u, is found using Equation 2.13 (Priestly and Park 1987 and Priestly et al. 1996). d u d y L p u y Lp h 2 (2.13) where, d y is the yield displacement, u is the ultimate curvature, y is the yield curvature, h is the height of the column, and L p is the length of the plastic hinge. The shear strength of a RC column, P s, is evaluated according to the following equations: P s S S (2.14) c s S c c c c b d (2.15) c e pt c S s A d w sy sin cos 1.15a (2.16) where, P s is the shear strength, S c and S s is the shear capacity resisted by concrete and transverse reinforcement, c is the average shear stress that can be borne by concrete, c c is the cyclic loading effect factor which can be obtained from Table 2.7, c e is the effective height factor, c pt is the modification factor depending on the longitudinal tensile reinforcement ratio, b is the width of the column section, h is the effective height of the column section, A w is the sectional area of reinforcing bars with interval a and angle and is the yield point of the reinforcements. sy 23

40 Table 2.7: The cyclic loading effect factor, k c. Load Type k c Static loading 1. Type I ground motion.6 Type II ground.8 motion The failure mode of a column is classified as either flexural failure, shear failure after flexural yielding, or shear failure and this is to be evaluated using Equation The failure mode is decided based on the ultimate lateral strength P u, shear strength P s, and shear strength under static loading P s of a RC column. s Pu P s : Flexural failure P P P : Shear failure after flexural yielding u s (2.17) Ps P u : Shear failure The lateral strength of the RC column, P a, is calculated depending on the failure mode using Equation 2.18: Pu P a Ps : Flexural failure + shear failure after flexural yielding (2.18) : Shear failure The ductility capacity of the RC column, failure mode: du d 1 a d y 1 y : Flexural failure a, is also calculated depending on the : Shear failure after flexural yielding + shear failure (2.19) where, d u and d y are the ultimate and yield displacement, is the safety factor which is determined based on the Seismic Performance Level and the type of ground motion. See Table 2.8 for the safety factors. Seismic Performance Level Table 2.8: Safety factor. Type I ground motions SPL SPL Type II ground motions 24