ANALYSIS AND REMEDIATION OF THE PINOPOLIS DAM ABSTRACT. During the course of routine review and analysis Santee Cooper, their consultants, and

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1 ANALYSIS AND REMEDIATION OF THE PINOPOLIS DAM Mark Carter P.E. 1 3 Steve Collins P.E., Ph.D. Ray Pinson E.I.T. 2 Paul Rizzo P.E., Ph.D. 4 ABSTRACT During the course of routine review and analysis Santee Cooper, their consultants, and the FERC determined that the Pinopolis Dams did not have an adequate margin of safety against failure during and immediately subsequent to a postulated recurrence of the famous 1886 Charleston Earthquake. Subsequent soil testing, computer stability analyses, research of remedial alternatives, and implementation of a pilot testing program led to a seismic mitigation plan that was engineered to include state of the art ground modification in conjunction with downstream berms. In all, the work consisted of constructing nearly 3,500 column elements requiring over 30,000 tons of rock, installing over 10 miles of perforated drainage pipe, and placing nearly 310,000 cubic yards of berm material over the top of the stone column areas at various locations along the 6-miles of dam. This paper summarizes the evaluation of project parameters such as seismic loading and soil properties, the various remedial alternatives considered, the engineering calculations associated with the selected remediation scheme, and the construction methods used during implementation of the Pinopolis East Dam Seismic Mitigation Project. INTRODUCTION The Federal Energy Regulatory Commission (FERC) mandates periodic inspections and review of design basis for all licensed hydroelectric projects in the United States. During the course of routine review and analysis Santee Cooper, their consultants, and the FERC determined that the Pinopolis Dams did not have an adequate margin of safety against failure during and immediately subsequent to a postulated recurrence of the famous 1886 Charleston Earthquake. The Pinopolis Dams essentially consist of three earthen embankments, named the West Dam, East Dam, and the East Dam Extension as shown in plan view on Figure 1. The West Dam and East Dam are separated by a concrete gravity section that includes a lock 1 Manager, General Construction, Santee Cooper, Moncks Corner, SC Principal Engineer, General Construction, Santee Cooper, Moncks Corner, SC Lead Engineer, Division of Dam Safety and Inspections, FERC Atlanta Regional Office, 3125 Presidential Parkway, Atlanta, GA Principal, Paul C. Rizzo & Assoc., Expo Mart Suite 270-E, Monroeville, PA 15146

2 to allow navigation between the Cooper River and the impounded Lake Moultrie. The SANTEE RIVER East Dam Extension East Dam West Dam COOPER RIVER Figure 1. Location of the Pinopolis Dams in Eastern South Carolina West Dam was remediated during the late 1980 s by constructing downstream bolster sections in those areas where the foundation sand was particularly loose. At these bolster sections, the loose sand was over-excavated and replaced as structural fill. The remediation of the East Dam and the East Dam Extension, which are founded on loose sands similar to the West Dam, are the subject of this paper. Initial stability analyses resulted in acceptable factors of safety for the structures under static conditions. However, factors of safety representing post-seismic stability were found to be less than 1.0 following dynamic loading with a 0.42g peak horizontal acceleration. Following several years of soil borings, laboratory tests, computer stability analyses, and research of remedial alternatives, a remediation program was developed which consisted of two main elements - (1) stone columns designed to relieve excess dynamic pore pressure and (2) a downstream berm of varying height over the stone columns. The following sections discuss the design and construction aspects for seismic mitigation of the Pinopolis East Dam and East Dam Extension.

3 GENERAL INFORMATION The Santee Cooper Hydroelectric Project was constructed from 1939 to 1942 in an attempt to alleviate the effects of the Great Depression on the poverty-stricken Southeast. Over 156,000 acres of bottomland were impounded to form Lakes Marion and Moultrie which are connected by a 6.5-mile canal. At the time of construction, Santee Cooper was boasted as being the largest land-clearing project on the North American subcontinent while nearly 40 miles of dams and dikes were constructed in order to generate hydroelectric power for the region. The Pinopolis Dams were constructed to impound the 60,000-acre water body, Lake Moultrie. The dams are compacted-fill earthen structures having a combined total length of 7.5 miles. They are located in the vicinity of Moncks Corner, South Carolina as indicated in Figure 1. The dams and powerhouse are approximately 35 miles upstream of Charleston, SC and the Atlantic Ocean. The Santee Cooper Project, which combined the Santee and Cooper Rivers to form the two connected lakes, is situated on the Atlantic Coastal Plain. The foundation soils in this part of the Coastal Plain consist generally of about 70 feet of sand, silty sand, and clay overlying the dense, somewhat cemented Cooper Formation. The Cooper Formation is the preferred founding medium for most foundations in the Charleston area. The dams were constructed 60 years ago over a layer of relatively low-blow-count natural material that is subject to liquefaction during a substantial seismic event. This layer is generally 15 to 20 feet below the ground surface at the toe of the dam and varies from 5 to 10 feet in thickness. The weak layer overlies the Cooper Formation as shown in Figure 2. The residual strength of this loose sand is very low, in the range of 100 psf, as determined from back-calculation of known liquefaction failures, both static and dynamic, as shown later in Figure 3c.

4 Figure 2. Typical Cross Section of the Pinopolis East Dam The stiff clay and clayey sands above the liquefied layers of loose sand are relatively dense and the dam embankment is a well-compacted, low-plasticity clay (CL). The maximum embankment height found near the powerhouse is approximately 60 feet above the downstream toe; however, the average embankment height over the entire 6-mile length is around 28 feet. The crest is approximately 20 feet wide with a typical top elevation of 88.0 MSL. The upstream face is covered with 10 inches of porous concrete for slope protection and is inclined at an average 2.8 horizontal (H):1 vertical (V). The downstream slope of the dams are grassed and inclined at an average 2.25H:1V. The normal pool lake elevation is 75.0 MSL, which equates to 13 foot of freeboard for the impoundment structures. The controlling earthquake in the area, and for a major portion of the East Coast of the United States, is a postulated recurrence of the 1886 Charleston Earthquake. The epicenter of this event is estimated to have been near Summerville, SC northwest of Charleston and the scene of numerous low magnitude events as many as two or three per year. The 1886 event was felt as far away as New York and Pittsburgh and caused extensive damage in the Charleston Area, including liquefaction. Paleoseismic investigations conducted over the past 20 years have yielded evidence of prior events large enough to cause liquefaction, with a recurrence interval estimated at approximately 450 years. The Modified Mercalli Intensity of the 1886 Event is estimated to be in the range of X near its epicenter. Moment magnitude estimates (there were no records) range from 6.9 to 7.6, but most professionals accept M 7.3 as being a reasonable estimate. A recurrence of this event near its epicenter is estimated to cause a peak ground acceleration at the East Dam of 0.42g. This estimate is the basis of the seismic design criteria adopted for this remediation project. Initial post-seismic stability analyses used the conventional and generally accepted effective stress approach with estimated excess pore pressures in the non-liquefied stiff clays of the foundation and embankment. This approach lead to factors of safety of approximately 0.6 to 0.7 at various cross sections. At this time (1994), Dr. Collins joined the regulatory oversight team and recommended modification of two important factors in the stability analyses, namely: Use of undrained shear strengths for the stiff, low permeablity clayey soils, For wedge or sliding block failure mechanisms, change from the Spencer solution method to the Lowe & Karafiath procedure, also commonly called the Corps of Engineers' method. This change was in solution techniques only, and did not incorporate the strength interpolation procedures ascribed to these methods.

5 The reasons for and implication to stability of these modifications are as follows: The first recommendation acknowledges that the two-step stability analysis developed by John Lowe et al, using undrained strengths for the embankment and stiff foundation strata, would better represent a potential dam slope failure in the moments and first several days following an earthquake at this particular site. In this short time frame, the low permeability soils must fail undrained, neither expelling nor taking up water. Migration of pore pressures from the deep liquefied stratum and from the reservoir would eventually eliminate the negative pore pressure effects that provide relatively high, undrained strengths in stiff soils. To prevent the migration tendency, a row of closely spaced drains at the downstream toe, that would bleed off the elevated pore pressures in the liquefied zone, was mandated as a minimal requirement whenever undrained strengths of non-liquefiable soils was used. Finite difference analyses concluded that the these liquefied strata pore pressures would be significantly reduced after two days by effective drainage into stone/sand columns at the toe. The second recommendation arose when noting that the inter-slice force for wedge failure surfaces was nearly horizontal when using the Spencer solution routine, implying no shear interaction between blocks. While the Spencer routine produced a lower FS for the critical failure surface compared to circular failure surfaces, the implication for stresses and forces is not kinematically possible. The critical failure surface is deep, passing through a long, relatively horizontal zone of liquefied sand, and then rising as a passive wedge to the ground surface, daylighting downstream of the toe of the dam. The upward movement of this wedge creates a shear displacement with the central block, and a downward shear force on the passive wedge at this interface, increasing stability in either drained or undrained soils. The fact that "complete equilibrium" programs satisfy equilibrium does not guarantee that the stresses are kinematically possible, or that the resulting FS is reasonable. The solution routine was switched to the methods developed by John Lowe, wherein the inter-slice angle could be pre-chosen. This value was conservatively set at 10 degrees, improving the stability results in both the passive and active wedge areas. The net effect of making these two modifications to evaluating the post-seismic stability resulted in a typical FS for the East Dam of around 1.10, below the required value of 1.3, but indicating much improved conditions when compared to the values obtained in the conventional analyses. A typical post-seismic FS for the East Dam Extension was 1.0, also indicating remediation was necessary. ADOPTION OF UNDRAINED STRENGTH PARAMETERS In order to perform the stability and deformation analyses, the shear strength for the soil predicted to undergo shear failure was assessed under three classifications: Non-Liquefied Soils;

6 Liquefied Soils; and Treated Soils. The shear strengths for Non-Liquefied Soils used in the stability analysis are summarized in Table 1 and typical summary plots of triaxial test data are shown in Figures 3a and 3b. SHEAR STRENGTH ON FAILURE PLANE (PSI) (PSF) STRENGTH ENVELOPE SATURATED EMBANKMENT 1,440 2,880 4,320 5,760 7,200 8,640 C = 864 PSF CS Strength Envelope based on regression analysis NORMAL STRESS (PSI) calculated used E4 E10 5,760 φ=25.8 4,320 2,880 63C CS Strength Envelope 67D used for Stability Analysis 1,440 69D φ=25.8 (PSF) Figure 3a. Soil Strength Plot for Saturated Embankment Material SHEAR STRENGTH ON FAILURE PLANE (PSI) (PSF) STRENGTH ENVELOPE STIFF CLAY (ALL DATA) 1,440 2,880 4,320 5,760 7,200 C = 1,034 PSF CS Strength Envelope based on regression analysis and used for Stability Analysis φ= NORMAL STRESS (PSI) 68C 67G E11 E25 64A 70C 8,640 4,320 2,880 1,440 (PSF)

7 Figure 3b. Soil Strength Plot for Stiff Clay Material Table 1. Strength Parameters (Non-Liquefied Soils) Drained Strength Undrained S trength* Material φ ( o ) C (PSF) φ ( o ) C (PSF) Unsaturated Embankment Saturated Embankment Stiff Clay Clayey Sand Ditch Fill Materials Rip-Rap Dense Sand Borrow Pit Mud SP material * C and φ refer to the intercept and slope in the τ ff versus σ fc format. A few supplementary comments pertaining to Table 1 are appropriate: 1. Undrained strengths were used in all instances where at least one row of stone columns, spaced 15-foot on centers, were to be installed at the toe of the dam, parallel to the axis. Utilization of undrained strengths for the stiff clay in the dam and foundation was a critical consideration in the stability analysis of the remediated structure. 2. Conservatively, the Upper Sands were assumed to liquefy unless the SPT blow counts were relatively high (N>30), ignoring any potential base isolation effects provided by the lower sands. In areas where these soils are near the surface at the toe of the dam, the lower of the drained strength or the post-liquefaction residual shear strength was used in the analysis. 3. Drained Strength Parameters were used for all SP soils with proper allowance for dynamic pore pressure. Figure 3c was used to assess the Residual Undrained Shear Strength of liquefied soils. This figure is taken from the work of Seed and Booker, 1976 and modified by expanding the abscissa to accommodate a broader (N1) 60-CS range. The undrained residual strength for the Lower Sand layer ranged between 75 psf and 435 psf depending on the (N1) 60-CS. The undrained residual strength for the Upper Sand Layers was interpreted to be in the range of 240 psf and 940 psf, which again depended on the (N1) 60-CS.

8 Figure 3c. Undrained Residual Shear Strength With the foundation and embankment soil strengths identified, Santee Cooper conducted a full-scale stone column test program to assess the technical and economic issues associated with the use of vibro-replacement as a remediation technique. Vibroreplacement, commonly referred to as stone columns, is generally accomplished by forcing rock into the ground by a combination of compressed air and vibration to form a dense column element. The goal of which creates a composite material of lower overall compressibility and higher shear strength than that of the native soil alone. Three separate test areas were selected along the toe of the East Dam and East Dam Extension. At each Test Area, nominal 42 diameter stone columns were installed in a triangular pattern with various spacings ranging from 6 to 10 feet. The evaluation was conducted by comparing Before and After Standard Penetration Test Results and Cone Penetrometer Test Results. In addition to verifying the applicability and cost issues associated with stone columns at this specific site, several important conclusions were developed for use in the final design. 1. Regardless of the soil classification, it was not practical to obtain adequate density to preclude liquefaction as measured with the SPT in the loose sand stratum at the

9 bottom of a stone column. Some other means of supplemental strengthening for this marginally improved material had to be considered. 2. The fines content of the soil being improved can have a significant negative effect, to the degree that, for some of the lower sands containing a relatively high percentage of fines, the desired benefits associated with stone columns are not realizable. 3. Using two sets of SPT Borings, we assessed the potential for beneficial aging in the treated soils. We found no real increase in N values due to aging within a time frame of about 4 to 6 weeks and again after four years with tests in Because of the inability to achieve substantial improved density at the bottom of the loose sand overlying the dense Cooper Formation, all parties concluded that the use of stone columns to preclude liquefaction could not be accomplished at this particular site. 5. Enlarged diameter bases could be achieved in the lower sands without excessive heave or other negative consequences to the existing dam. However, conducting the pilot test program was extremely beneficial, since it allowed Santee Cooper to eliminate a remedial alternative that was not viable for the project. It also allowed the designers to focus on the use of stone columns for drainage purposes and as structural elements, rather than densification, as a more efficient means of achieving increased shear resistance in the foundation. Therefore, the method of treatment for potentially liquefiable soils chosen for remediation of the East Dam and East Dam Extension consists of the installation of one to several rows of stone columns at the downstream toe in combination with an overlying berm where necessary. The installed columns supply the necessary drainage of earthquake-induced excess pore pressures and justify the use of undrained strength parameters in the stability analyses. Hence, the strength of the treated soils is based on the composite strength of the stone columns and the soil between the columns where multiple rows are used. SELECTING THE REMEDIAL APPROACH Throughout the data collection and analysis process, numerous options for seismic remediation were considered including the following: Over-excavation and replacement of the loose soil foundation layers at the toe of the dam, Various forms and types of grouting, Dynamic compaction with a falling weight, and

10 Downstream bolsters similar to the West Dam remediation, with stone columns in the foundation. With due consideration given to lessons learned from the West Dam project and from the pilot testing program, engineers selected stone columns used strictly as a structural and drainage element. Construction of an overlying berm would also be necessary at various sections along the dam to enhance the vertical stress in the columns, thereby increasing the post-seismic stability of the structures. Much effort was expended towards optimizing the remedial design by varying the column diameters, spacing, and berm height to target the most economical alternative. Each time a specific diameter, spacing, and berm height was selected, an individual stability run was necessary to ensure that the factor of safety and deformation criteria had been achieved. Designers and engineers had to find the break-even point of adding berm, increasing column diameter or rows, or varying the column spacing. These four (4) variables were rarely consistent for each cross-section analyzed due to the variability of foundation soils and the initial factor of safety at each specific location along the dam. For the East Dam, this typically required raising the FS from approximately 1.10 to The design process is generally termed the "Japanese" method, and details can be found in literature by Goughnour, et al. Once a stone column and berm configuration is chosen, the shearing strength through the column must be calculated, which depends on the shear surface inclination. For this project, the critical failure surface was known to be through the weak liquefied sand stratum. Due to its thin nature, only horizontal failure surfaces needed to be considered. The general scheme for the East Dam is shown in Figure 4. The equilateral triangular arrangement for stone columns was chosen, with two rows of enlarged base stone columns constructed on 12-foot centers, with bases 6 feet in diameter in the loose sand and nominal 30-inch columns rising to the surface. This configuration equates to one stone column per six linear feet of dam. The capacity of the stone column is thus divided by 6 feet to yield the applicable force in the slope stability program. This force per foot is further divided by a length representative of the stone column remediation area, in this case 20.8 feet, to yield a stress, or cohesion, for input into the stability program. The shear capacity of the column depends on how much load from the berm is attracted into the stiffer column versus the area subject to liquefaction between columns, and is expressed as a ratio S rv. Taking a simplified approach, which does not fully account for strain compatibility, Goughnour et al, recommend a range of 3 to 5 for S rv. A value of 3 is used in the example below. In this case, it is the higher modulus of the enlarged stone pedestal compared to the loose sand surrounding the pedestal that allows the column to attract more than the average vertical stress from the overlying berm. Thus, the column is effectively 6 feet in diameter rather than 30 inches.

11 Given: γ T Berm = 125 lb/ft 3 γ T Column = 110 lb/ft 3 Berm Height = 9.5 feet 30-inch columns w/ 72-inch enlarged base = 28.3 ft 2 2 rows of columns, spacing = 12 feet o.c. Calculation: Berm stress = 9.5 feet x k/ft 3 = 1.19 ksf average Foundation stress = 15 x ( ) = ksf Unit cell = 6 ft x 20.8 ft = ft 2 In a 6 ft x 20.8 ft cell (124.8 ft 2 ) there is one column: A.8 ft 2 r = area ratio = 28.3 ft 2 /124 = S rv = 3 = σ s /σ c σ c =.33 σ s σ s (0.227) +.33 σ s ( ) = 1.19 ksf σ s = 2.46 ksf Vertical stress on top of base (berm + foundation overburden): 2.46 ksf ksf = 3.17 ksf Shear from column: 3.17 ksf x 28.3 ft 2 x tan φ; where φ = 36Ε for # 789 granite stone = 65.2 kips Shear from liquefied soil around column, with residual strength of 100 psf: ( ) x (124.8 ft 2 ) x 0.1 ksf = 9.6 kips Total shear : 65.2 kips kips = 74.8 kips Stress: 74.8 kips/124.8 ft 2 = 0.60 ksf, treated as cohesion This 0.60 ksf is then treated as cohesion in a zone 20.8 feet wide in the loose sand stratum, representative of the conditions at the top of the loose sand. Minor increases in strength with depth in the stratum are neglected. Corresponding calculations up the 30- inch column could be made, but at this site the additional strength provided in the stiff portion of the foundation was not needed to produce acceptable FS values.

12 Computer runs of slope stability were then examined to confirm that the critical failure surface was horizontal through the remediation area. The S rv of 3 and φ = 36Ε for the stone column base are lower limit values that provide the additional force to raise the stability of the critical failure surface to FS = Values in the range of S rv = 4 and φ = 42Ε would have been acceptable, yielding a cohesion of 0.80 ksf and somewhat higher calculated FS. Recognition of a modest N value increase of 3-5 in this remediation zone, as determined in the 1995 test sections, would also increase the strength by an additional 50 psf, and has been neglected in the above calculations. The adopted shear strength took no credit for dissipation of pore pressures during or after the earthquake. Post-seismic stability must be maintained immediately after the earthquake when the dynamic excess pore pressures are still in effect and not yet dissipated. As discussed previously, the dam stability will increase with time as the pore pressures dissipate in the liquefied zone. Within several days, the liquefied zone will reconsolidate to the point where the pre-earthquake static stability has been reestablished. CONDUCTING THE REMEDIAL WORK In January of 2002, Santee Cooper initiated the stone column portion of the Pinopolis East Dam and East Dam Extension Seismic Mitigation project. The specialty contractor performing the work was Hayward Baker, Inc. (HBI), of Tampa, Florida - specializing in ground modification, site improvement, and soil stabilization. Installation of the columns was accomplished by using the Dry Bottom-Feed Vibro-Replacement method in accordance with the FERC-approved remediation plan. The rock backfill used for the stone column work was SCDOT #789 granite stone. Extensive filter criteria analyses were conducted to verify the adequacy of the 789 stone for column production in the Pinopolis Dam foundation soils. Several grain size distributions representative of the liquefiable layers along the dams were compared to the gradation of the 789 stone. Typical filter criteria comparing the D 15 of the filter material to the minimum D 85 of the contact soil were utilized. The 789 stone met this criterion and was selected for use on the project. A supplier for the aggregate was found to be only two hours from the project site. In order to allow for penetration of the vibrator, the Contractor pre-augured each stone column using a 30-inch diameter auger as shown in the Figure 5. The 30-inch diameter hole was predrilled one foot into the dense Cooper Formation, which was generally 30 feet below the ground surface. After reaching full depth, the auger was reversed to reduce the amount of excess material removed from the hole. The 72-inch diameter column portion was installed starting approximately one foot into the dense sand and the column was then constructed in approximate 2-foot lifts. Pursuant to the project specifications, when the recorded vibrator amperage exceeded 70 amps of the free hanging amperage, the minimum 72-inch diameter requirement was waived and the crane operator pulled up to the next lift. As expected, this scenario rarely occurred. Next, the

13 30-inch diameter column portion was installed from the bottom of the stiff clay layer to the ground surface. Figure 5. Drill Rig and Stone Column Crane at the toe of the Pinopolis East Dam Stone Column Construction Procedures: The Contractor installed the stone columns utilizing a nominal 30-inch diameter bottom feed S-23 vibrator. The vibrator was driven by an electric motor powered by a 250 kw Caterpillar diesel generator. The motor type was a 3-phase squirrel cage, delivering 161 hp at 1,800 rpm, producing 23 tons of centrifugal force. The length of the base vibrator was approximately ten feet and the weight was three tons. Extension tubes were added to reach the treatment depths. The stone backfill was delivered to the tip of the vibrator using an eightinch diameter feed tube attached to the vibrator. The top of the feed tube was connected to a pressure chamber with a 1.25 cubic yard capacity to hold the stone for delivery to the vibrator tip. With the chamber pressurized, an air-controlled valve operated the door to allow the rock to fall into the chamber from the hopper. The stone was delivered to the hopper via a two cubic yard skip that was raised to the hopper once it was filled with stone from a front-end loader. The entire vibrator assembly was suspended from a 150-ton

14 crawler-type crane as indicated in Figure 6. Figure 6. Stone Column Crane and Vibrator The stone column production was monitored by the Contractor using a real-time automated recorder. The data recorded included the column identification number, installation time, depth of probe, height of each lift, weight of material added to each lift, column diameter for each lift, and air pressure of the gravel feed system. This data was provided to the project engineer in the form of daily reports that included a log of pertinent information and a graph for each column. Each graph showed the column diameter, column depth, construction amps, penetration amps, and air pressure. The daily logs were used to tabulate the cumulative column depths and rock quantities that were contract pay items. Also, each graph was used for quality control purposes since the minimum column diameter and vibrator amperage had to be achieved with very little tolerance pursuant to the project specifications. The following work was accomplished over a 10-month period ending October 4, 2002: Two (2) rows of 30-inch top diameter with 72-inch base diameter stone columns were installed on 4,125 linear feet of the East Dam proper. HBI constructed 553 stone columns with a cumulative footage of 16,251 feet requiring over 8,800 tons of rock backfill. On the East Dam Extension, three (3) rows of 30-inch diameter stone columns were installed on 9,300 linear feet and a single row of 30-inch diameter columns was installed in a 9,000 linear foot section of dam that required pore pressure relief only. HBI constructed 2, inch diameter stone columns on the East Dam Extension with a cumulative footage of 98,103 feet requiring over 21,000 tons of rock backfill. Installation of the Drainage Netwo rk: Following installation of the columns, consideration had to be given to conveying water from the columns that would be generated from relief of excess pore pressures during and after a seismic event. As opposed to a conventional sand/gravel drainage blanket, a piping network was designed as a more economical alternative. As indicated in Figure 7, the piping system is composed of 18 wide

15 Figure 7. Placement of Drainage Pipe at Toe of Dam corrugated and perforated high density polyethylene pipe to route upward flow from the columns and discharge the accumulated water out the toe of the new berm. The pipe is offset from the high vertical stress zone over the columns, and connected by a shallow trench backfilled with 789 stone. A six foot wide strip of filter fabric over the columns and trench prevent infiltration of soil fines from the overlying berm. The flow capacity of this pipe used in a horizontal manner with 1 of head is 28.0 gpm. Flow rates from the columns using relief well calculations and surrounding soil permeability were conservatively estimated to be 1.5 gpm. The piping network shown in the photograph parallels the dam and continues through the areas where multiple rows of columns were installed with an overlying berm. This drainage design met the projected flow requirement and represents distinct advantages from an installation aspect in terms of uniformity, handling, and cost. Berm Construction: As mentioned in the remedial approach section, the columns needed to be loaded with mass in order to effectively increase the composite soil strength at depth. Therefore, a berm ranging in height from 9.5 to 13.5 feet was constructed over the stone columns. Various berm crest widths, ranging from 56 to 40 feet, were required to cover the different spacings and rows of stone columns. On the East Dam there was sufficient area available to flatten the berm slope to 2.5H:1V; however, the East Dam Extension had very little room to work as a result of the original borrow pit location. Since original construction in 1940, the borrow pit area has evolved into a pristine wetland area providing habitat to many species of plants and animals. Therefore, the engineering design in this area had to take into account the potential environmental impacts associated with the work. Tighter column spacings and a 1H:1V berm slope prevented encroachment into the downstream wetland boundary. The materials for berm construction were obtained from borrow areas in the vicinity of the project and meet the USCS classification of SW, SM, SP, SC or CL. The berm material was specified to have no more than 50% fines (material passing the #200 sieve) by weight. Over 300,000 yards of berm material were placed in compacted lifts of 6 inches or less and compacted to 95% of the maximum dry density obtained from the Standard Proctor test (ASTM D-698). The moisture content was required to be + 4% of the optimum. A combination of smooth drum and padfoot vibratory rollers were used to accomplish compaction which was verified by field testing in accordance with ASTM D and ASTM D SUMMARY The Pinopolis East Dam Seismic Mitigation Project was a cost effective solution to a seismic deficiency problem. With a combination of strategically placed stone columns

16 and downstream berm material, Santee Cooper was able to meet the current earthquake standards for dam structures in addition to safeguarding the downstream floodplain, which includes portions of metropolitan Charleston, SC. The final plan also considered space constraints due to the presence of jurisdictional wetlands downstream of the dam and the economies of scale associated with berm height versus diameter and spacing of stone columns. Several valuable lessons were learned during the analysis, design, and construction phases of the project as follows: In certain circumstances, the use of undrained soil strengths for non-liquefied, stiff clayey soils can be appropriate in evaluating remedial alternatives for embankment stability. Engineering a viable solution also requires a substantial amount of economic analysis. Engineers and designers expended much effort on optimizing the remedial design by varying the column diameters, spacing, and berm height to target the most economical alternative. When applicable, implementation of a pilot testing program can confirm whether or not a selected remedial method will achieve the desired results. The stone column full-scale testing program allowed engineers and designers to evaluate a remedial alternative that is typically used for in-situ soil densification. Test results showed that the stress transfer method was a better technique for the specific site conditions and presented a viable option for increasing the factor of safety against dam failure. During the past 20 years, Santee Cooper has completed an extensive program of seismic analysis, design, and construction for its project embankments. At the conclusion of the Pinopolis East Dam work, Santee C ooper achieved the dam safety goal of meeting the current seismic evaluation standards for its project structures. REFERENCES Barksdale, R. D. and Bachus, R. C., Design and Construction of Stone Columns, Report FHWA/RD-83/026, FHWA, Goughnour, R. R., Sung, J. T., and Ramsey, J. S., Slide Correction by Stone Columns, Deep Foundation Improvements: Design, Construction, and Testing, ASTM STP 1089, Melvin I. Esrig and Robert C. Bachus, Eds., American Society for Testing and Materials, Philadelphia, Seed, H. B. and John R. Booker, Stabilization of Potentially Liquefiable Sand Deposits Using Gravel Drain Systems, Report No. EERC 76-10, Earthquake Engineering Research Center, University of California, Berkeley, California, 1976.

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